del Cemento
The 3rd International fib Congress
Washington DC, May 29 – June 2, 2010
Italian National Report
ISSN 0019-7637
ISSUE € 30,00
Research and Construction
Anno/Year LXXX
Aprile/April 2010
Direttore responsabile
Managing Editor
Laura Negri
l’industria italiana
del Cemento
3 Foreword (M. Menegotto)
Marco Veronesi
Grafica e Impaginazione
Design & Editing
Studio Mariano - Roma
Earthquake Engineering of Reinforced Concrete Structures: The Italian
L. Ascione, E. Cosenza, G. Mancini, G. Manfredi, G. Monti, P. E. Pinto
Experimental Research on Seismic Behavior of Precast Structures
F. Biondini, G. Toniolo
Direzione e redazione:
Piazza Guglielmo Marconi, 25 - 00144 Roma - Tel. 06/54210237 Telefax 06/5915408 - E-mail: [email protected]
Autorizzazione del Tribunale di Roma n. 301 del 24 Ottobre 1950.
Concessionaria per la pubblicità:
Idra S.A.
Strada Cardio, 4 - 47891 Dogana (RSM)
Tel. 0549-909090, fax 0549-909096
e-mail: [email protected] – www.idrabeton.com
Damages of L’Aquila earthquake
G. Manfredi
Reconstruction between temporary and definitive: the CASE project
G.M. Calvi, V. Spaziante
PUBBLICEMENTO s.r.l. - Sede legale:Viale Ettore Franceschini, 37 00155 Roma. Sede amministrativa e operativa:
Piazza Guglielmo Marconi, 25 - 00144 Roma
Tel. 06/54210237 - Telefax 06/5915408
Fascicolo/Issue 30 €
Tutti i diritti di riproduzione sono riservati. Nessuna parte di questa
Rivista può essere riprodotta in nessuna forma. La Rivista non
assume la responsabilità delle tesi sostenute dagli Autori e delle
attribuzioni relative alla partecipazione nella progettazione ed
esecuzione delle opere segnalate dagli stessi Autori.
All rights of reproduction are reserved. No part of this Magazine may be
reproduced in any form whatsoever.The Magazine assumes no
responsibility for the theses put forward by the authors or for the authors'
indications of their responsibilities regarding participation in the design
and construction of the works.
Associata all’Unione Stampa Periodica Italiana.
ISSN 0019-7637
Stampa/Printed by: Grafica Ripoli snc - Via Paterno - Villa Adriana (Tivoli)
In copertina
On front cover
Pantheon dome, Rome. Inside view
State-of-the-art on Research on Structural Concrete in Italy
M. Collepardi
Italian High-Speed Network. A special focus on concrete structures
• HS railway cable-stayed bridge over Po river
• “Piacenza” viaduct
• “Modena” system viaducts
• “Savena” viaduct
• “Caivano” variation structures
• Tunnels in the Florence-Bologna stretch of High-Speed Line
• New stations for Italian High-Speed Network
“Colletta” cable-stayed bridge
Rio S’Adde viaduct
Bridge over Vajont creek
“Cesare Cantù” cable-stayed bridge
Bridge between La Maddalena and Caprera Islands
“Don Bosco” bridge. Architecture, white as light
Viaduct for State Road (SS) 23
“Roccaprebalza” viaduct
“Sandro Pertini” bridge upgrade
Cable-stayed footbridge over the Frodolfo river
Bridge over Mazzocco creek
Bridge over the Sacco river
Bridge over the Santa Caterina channel
Bridge over the Cimadolmo branch
“Isola della Scala” bridge
The “Strada dei Parchi”
A24 – Completion of the motorway Roma-L’Aquila-Teramo
“S. Antonio” viaduct
Viaduct for the Algeria East-West Motorway
l’industria italiana
del Cemento
MAXXI – Center for the contemporary arts
“Olympic Palavela”
New Bocconi University
Milanofiori 2000 – Corporate Center
“Acqua minerale San Benedetto” plant
New Sky Italia Headquarters
Light Pavilion
Agenzia Spaziale Italiana new headquarters
“Altra Sede” for the Regione Lombardia
New “Sant’Anna” Hospital
“Verdi” Theatre
Banca Lombarda Center
Boglietti Palace
“Cuore immacolato di Maria” parish complex
“San Giovanni Battista” parish complex
“Somada” Business Center
Edited by A.I.C.A.P. (Associazione Italiana Calcestruzzo Armato Precompresso) and “l’industria italiana del Cemento – ii C”
The 3rd International fib Congress
Italian National Report
Research and Construction
he 3rd fib International Congress takes place during (hopefully towards the end of) a global downturn in the economy.
The impact was felt in all fields, including construction industry, and in all countries.
Crises are opportunities for rethinking activities and habits. The need to face high costs of materials and production stimulates
the search for improvements. This parallels the awareness of further needs, not of immediate return, such as saving resources,
building more durable constructions, recover for reuse or recycling at least of materials, or better yet of whole structures or parts
of them.
Design and construction usually take account of safety and economy but more and more they consider too service life
requirements, products’ life cycle assessment, resource saving and other environmental issues, in one word, sustainable
CEB and FIP, now fib, were accompanying and leading the progress of structural concrete design and construction during the
past half century and have been a great partner in cultural interchange for our country.
AICAP, the national association for structural concrete, mirrors fib and takes advantage also of its actions to contribute in its
turn to the dissemination of knowledge and the improvement of practice. Among other initiatives directed toward encouraging
engagement in better design and execution, AICAP launched for the first time a national award for the best concrete structures
for buildings and civil engineering works, that will be given every second year, and edits this National Report, where a
selection of works using structural concrete and completed in Italy in the last four years is illustrated, among which both 2009
award winners. Rome’s Pantheon is the exception, with its record 43.4 m dome shown on the cover, cast altogether 20 centuries
ago, using a lightweight concrete quite similar to the modern one.
Research is playing a decisive role in advancing techniques to fulfil new requirements and has progressed in Italy during these
years, promoted also by the industry, more sensitive to it while under pressure of the crisis. Therefore, this Report also describes
some results related to the national state of the art.
Italy is a highly seismic country, thus most researchers in recent past have worked in the field of seismic engineering, aiming at
better design of new structures as well as at assessing and retrofitting older ones.
Unfortunately, one year ago this science was called to prove its ability after the L’Aquila earthquake. The authors of the articles
in this Report were personally involved also in the field operations. Their contributions helped the surveys in the immediate
post-event days, the directives for the repairs and the implementation of the CASE project. The latter has represented quite a
successful system response, up-to-date and praised worldwide, that provided a large population of homeless with sets of
permanent houses, built in few months on seismically isolated platforms, located in sites purposely selected and equipped in the
city neighbourhoods.
AICAP is then glad to present the 2010 National Report, which comes in to being too thanks to the renowned magazine ii C,
which is publishing this special issue, in spite of the difficult times.
Marco Menegotto
Head of the Italian delegation to fib
Earthquake Engineering of Reinforced
Concrete Structures:
the Italian State-of-the-art
his extended paper summarizes part of the results, those related to
reinforced concrete (RC) structures, of the largest research program
on earthquake engineering ever held in Italy. The ReLUIS project funded by the Department of the National Civil Protection granted
15.000.000 € and involved more than 600 researchers all over the country between 2005 – 2008. The ten research tasks (lines) ranged from the
seismic risk of existing structures to new design paradigms, including
geotechnical earthquake engineering issues and innovative approaches to
seismic risk reduction as earthquake early warning systems as well as
emergency management.
In the following the ReLUIS research lines regarding: (I) assessment of
existing RC buildings (Line 2); (II) bridges (Line 3); and (III) retrofit of
RC structures via innovative materials (Line 8), coordinated by the
authors, are described in their development and findings. The three section of the paper are structured as stand-alone, including their own introduction, conclusions, vision and references, for readability purposes. More
details, references and products of the project may be found in the
research section of the ReLUIS website (http://www.reluis.it/).
Because of timeliness of the ReLUIS project and the involved scientists,
the work described in the following is likely to express the state of the art
of earthquake engineering concerning reinforced concrete structures in
Research Line 2 focuses on the assessment of the seismic performance
of existing reinforced concrete buildings, covering a wide spectrum of
problems, each one treated within a single Task. These aspects span
from those related to the preliminary knowledge phase, to the use of
nonlinear assessment methods, while placing emphasis on peculiar
modeling problems, such as those related to the presence of stairs,
infills, beam-column joints, and biaxial behavior of the elements. The
Research Line is also devoted to the study of mixed-type (masonry/RC)
buildings and of prefabricated industrial buildings.
The following Task list collects and organizes the entire scientific activity:
1–MND: Non-Destructive Methods for the Knowledge of Existing
Calibration of Confidence Factors
3–IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
4–MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
5–TAMP: Influence of Infills on Structural Response
Behavior and Strengthening of Stairs
7–NODI: Behavior and Strengthening of Beam-Column Joints
8–BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
9–PREFAB: Behavior and Strengthening of Prefabricated Industrial
For the sake of clarity of exposition, given the large variety of different
subjects, the structure of the paper, in each of its sections, follows the
Task organization above.
1.1 MND: Non-Destructive Methods for the Knowledge of Existing
Task MND specifically focuses on the knowledge of the constituent
material properties of Reinforced Concrete (RC) existing structures.
Particularly, Task MND is devoted to the estimation of the in-situ concrete strength by using destructive and non destructive methods. Data
on material properties from several in-situ and laboratory investigations
were collected and analysed with the major objective of defining reliable as well as not very expensive procedures and criteria for the estimation of the in-situ concrete strength. Further, methods for the treatment of the uncertainty that characterizes experimental data obtained
through in-situ and laboratory investigations were analysed and some
theoretical simulations of the influence of material properties on the
seismic capacity of existing buildings were carried out.
1.2 FC: Calibration of Confidence Factors
A fundamental phase in the assessment of existing reinforced concrete
buildings and in their strengthening design is the knowledge process
that one has to follow to acquire the necessary information. This is
based on the collection of different kinds of information regarding: a)
the structural system configuration, b) the materials strength, c) the
reinforcing steel details, and d) the conditions of the structural elements.
The Italian Code (OPCM 3431, 03-05-05, Annex 2) as well as the most
advanced International Codes (FEMA 356, EC8 Part 3) specifies data
collection procedures about the configuration of the structural system,
as well as material strength and condition of the structural elements
comprising the building, and ensuing Confidence Factors (CF) to apply
to the mean values materials properties, based on the quantity and
RESEARCH - Seismic behavior
Luigi Ascione1, Edoardo Cosenza2, Giuseppe Mancini3, Gaetano Manfredi2, Giorgio Monti4 and Paolo E. Pinto4
Università degli Studi di Salerno, Fisciano (SA), Italy
degli Studi “Federico II”, Naples, Italy
Politecnico di Torino, Turin, Italy
4 Sapienza Università di Roma, Rome, Italy
2 Università
quality of the information gathered (the so called Knowledge Level). In
the current approach, the CFs are given through tables.
Aim of the Task has been: a) evaluation of CF effects on the assessment
of buildings seismic performances; b) development of a procedure for
the evaluation of concrete and steel strength, to be reliably used in
assessing members capacity; c) new definition of CF, evaluated by a
closed-form equation as a function of number, kind and reliability of
each testing method employed, and of the reliability of prior information.
1.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
Task IRREG deals with problems related to the definition of plan and
elevation irregularity and the effects of irregularity on the structural
behavior and its prediction though different methods of analysis provided by the current design codes.
More specifically, the main focuses of this tasks are:
a. Study of the definition and effects of plan irregularity on the response
of RC buildings up to the Ultimate and Collapse limit states;
b. Comparison and calibration of different linear and nonlinear methods for reinforced concrete structural members, with emphasis on
pushover and nonlinear dynamic analyses;
c. Comparison between research-oriented and professional-oriented
structural analysis software, in order to identify analytical tools that satisfy both modeling precision and computational speed.
1.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
The work has been developed on the evaluation of the seismic response
of mixed-type buildings behaving as parallel systems, with regard to
both local (interaction between masonry and RC elements) and global
features, by performing a series of non linear numerical analyses.
The research activity has been focused on:
a. the classification of the main geometrical characteristics of such kind
of buildings and the study of their response – behaving as parallel systems – subjected to horizontal forces;
b. the problems concerning the modelling of mixed-type buildings and
on the distribution of the seismic action between masonry and reinforced concrete elements by performing a series of numerical analyses
obtaining the capacity curves of individuals resistance elements and
the building as a whole.
1.5 TAMP: Influence of Infills on Structural Response
Several theoretical and numerical analyses, and, above all, the damage
distributions on buildings that have suffered an earthquake show that
RESEARCH - Seismic behavior
masonry infills can modify substantially the expected seismic response
of framed structures although special devices connecting the infill panels with the surrounding meshes of frame are not applied.
In spite of that, most seismic codes (the more recent too) give some provisions in order that the resisting elements of frame bear the
unfavourable effects of a non-uniform infill distribution in plan or/and
in elevation, but they do not suggest any procedure to quantify these
effects or the favourable lateral stiffness and resistance contributions
that infills give when they are uniformly located.
This gap occurs since the influence of the masonry infills on the seismic response of framed buildings is a still open research topic, where
univocal and general results do have not been achieved. The present
study refers to this subject with the following main objectives:
- mechanical characterization of the masonry infill kinds that are commonly utilized in the Italian country by means of experimental tests on
their components (resisting elements and mortar) and masonry samples;
- experimental investigation on infilled meshes of RC frames, with the
aim of calibrating a pin-jointed equivalent diagonal strut model.
A companion study has also been devoted to verifying the influence of
the infills on the seismic response of RC framed structures; for this purpose, shaking table tests on a 1:2 scaled 3D building and numerical
nonlinear analyses of multi-storey frames have been carried out.
1.6 SCALE: Behavior and Strengthening of Stairs
The main objectives of this Task are the following: Identification of the
main stairs typologies used in the past construction practices;
Numerical investigation of the influence of the stair substructure on the
structural seismic response. In particular, both global and local seismic
performance have to be investigated with reference to frame and stairs
members connections; Construction of building sub-assemblages,
including a stair substructure, for experimental tests execution, specifically targeted at understanding their seismic performance.
1.7 NODI: Behavior and Strengthening of Beam-Column Joints
This Task aims at investigating the experimental behaviour of RC structural members, particularly beam-column joints without or with
strengthening, thus providing a contribution to a more reliable evaluation of the seismic vulnerability of RC existing buildings. In particular
of great interest is the understanding and the validation of capacity
models relevant to the joint panel zone in beam-column sub-assemblages reported in the literature and in seismic codes. Further, there is
a need of knowledge in the field of strengthening and retrofit systems
that can be used taking into account the actual geometry of joints: e.g.
presence of slab and other framing elements that could prevent an
effective arrangement of the retrofitting system. To this purposes, wide
bibliographic research on the experimental investigations on beam-column joints and on different repairing/strengthening techniques as well
as experimental researches on different joint specimens have been carried out.
1.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
Modern approach to safety assessment of existing reinforced concrete
structures and design of strengthening interventions, in particular those
aimed at increasing ductility of columns, are based on enhanced and
complex methods for structural analysis (seismic demand), but also on
the availability of data concerning performances of members at failure
(seismic capacity).
On the other hand, common constructions are not necessarily affected
by regular shapes and/or regular distribution of seismic resistant substructures, so that seismic actions result in complex deformation paths
on columns and in general on compressed resisting members.
This is the reason why Task BIAX research activity has been devoted
to provide an insight on the response of r.c. members subjected to biaxial bending and axial load. In particular, some aspects have been
analysed in detail. In compliance with the overall objectives of the
research programme as a whole, Task BIAX duties were the definition
of a set of reliable and well documented data and procedures concerning: (a) rotation capacity of r.c. members subjected to generalised bending and axial forces; (b) development of simplified methods of analysis
for general r.c. cross sections for design safety checks; (c) development
of refined methods for assessment of generalized moment-curvature
relationships of cross sections; (d) extension of results to r.c. members
reinforced with FRP materials.
1.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
The assessment and reduction of seismic vulnerability of a widespread
category of precast structures typically used for industrial buildings is
a topic of high importance. The production of these structures starts
since from the years ‘50s of last century with elements and construction
solutions which had a relevant evolution through the subsequent times.
It is a social important interest to know the state of this wide building
heritage with respect to its seismic vulnerability so to address, following rational criteria, possible interventions of upgrading of inadequate
The essential motivation for each Task stems from recognizing some
gaps in the code, related to certain procedural and methodological
aspects in the seismic assessment of existing buildings.
Specifically, for Task MND, it is noted that, in the case of non-destructive testing, a lack of clearness exists about the relative importance of
such tests with respect to destructive ones for the evaluation of material properties.
The data acquisition modality has immediate consequences on the calibration of Confidence Factors, treated in Task FC, which may assume
different values from those given in the code, in case one accepted to
include results from non-destructive tests in addition to – or even in
substitution of – destructive ones.
Moving to the level of analysis methods for seismic safety evaluation,
the need of a deeper insight into the usual assessment techniques is
recognized, with particular emphasis to their predictive capacity when
dealing with irregular buildings, dealt with in Task IRREG. It would be
expedient to identify, for example, a synthetic parameter capable of
quantifying the level of irregularity and, possibly, an associated applicability threshold that helped selecting the most appropriate assessment
method, be either of simplified nature, such as pushover analyses, or
more refined, such as nonlinear dynamic analyses.
For mixed-type (masonry/RC) buildings, the lack of code provisions,
which could guide the designer towards the assessment of the compound behavior in a unitary manner – also accounting for interface
actions between different constructive typologies – , is absolutely striking and appropriate methods and provisions should be identified in
Task MIX.
A different remark is needed for the influence of infills on the structural response, where the motivation for the research carried out by Task
TAMP stems from the absence of code provisions to account for the
meaningful interactions that develop between infills and structure, with
significant effects, both, at the global level (behavior factors), and at the
local level (collapse mechanisms induced by the presence of localized
The following three Tasks SCALE, NODI and BIAX, related to stairs,
beam-column joints, and biaxial behavior of columns, respectively, deal
with three aspects where the need of providing the designers with operational tools is imperative, especially for as regards the assessment of
the capacity of such elements. In the first case, the motivation is to
obtain a deeper insight about the influence of stiffening elements – the
stairs – on the structural response. Generally, when modeling the resisting system, these elements are either neglected or modeled with unacceptable simplifications. In the second case, that relative to beam-column joints, it is necessary to develop more accurate capacity models,
accounting for the joint panel behavior, but also for the presence of secondary phenomena significantly modifying the resisting mechanism,
RESEARCH - Seismic behavior
such as concentrated forces ensuing from hook-bent bars, or bond-slip
in rebars. In the third case, that of biaxial behavior of columns, the
motivation for research stems from the awareness that the capacity
equations currently available in the code are calibrated on the monoaxial behavior, besides, without interaction with shear.
Finally, for the prefabricated structures studied in Task PREFAB, the
intention is to provide the normative framework with more complete
indications than those currently available, with the objective of bridging the current information gap through the proposition of specific
guidelines for the seismic assessment and strengthening of such buildings.
The above considerations are expanded in the following sections.
2.1 MND: Non-Destructive Methods for the Knowledge of Existing Structures
Modern seismic codes require that a knowledge level (KL) is defined
(e.g. 3 KLs in EC8 part 3: limited, normal and full knowledge) in order
to choose the admissible type of analysis and the appropriate confidence factor values in the evaluation. Among the factors determining
the KL, there are the mechanical properties of the structural materials.
In RC structures, the compressive strength of concrete has a crucial role
on the seismic performance and is usually difficult and expensive to
estimate. Reliable procedures to take into account the factors influencing the estimation of in-situ concrete strength, particularly in case of
poor quality concrete, are not currently available. According to various
codes (e.g. in Europe EC8-3, in Italy NTC 2008) estimation of the insitu strength has to be mainly based on cores drilled from the structure.
However, non-destructive tests (NDTs) can effectively supplement coring thus permitting more economical and representative evaluation of
the concrete properties throughout the whole structure under examination. The critical step is to establish reliable relationships between NDT
results and concrete strength. The approach suggested in most codes
(e.g. in EC8-3) is to correlate the results of in-situ NDTs carried out at
selected locations with the strength of corresponding cores. Thus, NDTs
can strongly reduce the total amount of coring needed to evaluate the
concrete strength in an entire structure.
2.2 FC: Calibration of Confidence Factors
Data collected for the assessment of a building are obtained from available drawings, specifications, and other documents for the existing construction, and must be supplemented and verified by on-site investigations, including destructive and nondestructive examination and testing
of building materials and components.
As a function of the completeness of as-built information on buildings
(Knowledge Level) the Italian Code specifies different analysis methods
and Confidence Factors (CF) to be applied to the mean values of mateRESEARCH - Seismic behavior
rials strengths.
Difference in the knowledge procedure about the single structural parameters and the actual possibility of propagation to the structure as a
whole of information gathered on single members unlikely can be
accounted for by a single CF to be applied to mean materials strength
Material strength is characterized by, both, an intrinsic spatial variability and an epistemic uncertainty, caused by either workmanship (for
instance not compliance with the original project, execution of structural elements in different times with different materials strength), or
reliability of testing methods, or degradation of material properties with
time, or a combination of the former. On the other hand, amount and
detailing of reinforcement, defective detailing, etc., neglecting the
intrinsic uncertainties, are characterized by epistemic uncertainties
only, mainly due to lack of the original project and/or not compliance
with it; collected data on one structural element are certain but do not
allow to eliminate uncertainties about other elements.
Objectives of recent studies (Franchin et al. 2008, Jalayer et al. 2007)
have been the evaluation of the effect of CF on the assessment of the
structural reliability and new proposals for calibration of a CF.
2.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
Many of the existing RC structures were built without accounting for
seismic actions, thus much attention has been paid in recent years to
the development of reliable methods of analysis and assessment. Linear
methods seem inappropriate in most cases; many current seismic codes
and guidelines include provisions for nonlinear analysis (Eurocode 8,
2003a, EuroCode 8, 2003b, FEMA 356, 2000, ATC-40, 1996), which
seems to be the natural choice for existing structures subjected to moderate and strong design earthquakes. This is obviously a big issue in
Italy, a seismically active country where many buildings were erected
in the ‘60s, ‘70s and ‘80s usually accounting for only gravitational
actions. Furthermore, the new seismic zonation classifies areas previously considered non-seismic as seismic, thus new assessment are
needed even on recently built structures.
Following the publication of the most recent Italian Seismic Codes, the
ReLUIS program of the Italian Department of Civil Protection intends
to validate and improve the new code, to propose alternate procedures
when deemed necessary, and to provide practical examples to practicing engineers. These activities are particularly important for new
methodologies, such as nonlinear methods of analysis. Focus of these
studies is not only the application of the nonlinear methods of analysis,
but also the use of the results of the nonlinear analyses to assess the
seismic vulnerability of structure.
2.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
From the early 20th-century the combined RC-masonry buildings widely spread in European, Mediterranean and Southern America countries.
Despite the diffusion of this combined building typology, the international guidelines have not followed building evolutions; nowadays,
international guidelines are not exhaustive to deal with specific problems of this building typology, such as: horizontal loads repartition, connections between different technology elements and over strength factor.
The Argentinean guideline (NAA-80) points out the fundamental role
performed by slab, on the base of the own relative stiffness, for sharing
seismic action between vertical different technology resistant elements.
During the years, the Italian guidelines have provided discordant indications.
The Italian guideline (D.M. 1996) suggested to assign the total seismic
action to masonry walls in the case of new buildings, while for existing
buildings, the combined RC-masonry buildings should be considered
as structural elements typology that prevalently supports horizontal
loads, generally masonry walls.
Regarding masonry buildings, the Italian guideline (O.P.C.M. 3431)
allows to employ different technology elements to support gravity loads,
only if the seismic action is fully supported by elements with the same
technology. In the need to consider the collaboration of masonry walls
and different technology systems to sustain the seismic action, a nonlinear analysis should be carried out according to O.P.C.M. 3431. The
latest Italian code (D.M. 2008) confirms the instructions provided by
O.P.C.M. 3431 by which the real structural system should be considered with particular attention to, both, stiffness and strength of the
slabs, and the connections effectiveness between the structural elements.
2.5 TAMP: Influence of Infills on Structural Response
The very numerous papers that concern the behaviour of infilled frames
are quite uniformly distributed within the last forty years (Figure 1a).
This subject has kept topical mainly because of the following reasons:
- different materials that can be utilized for the infill panels; - difficulty in modelling the frame-infill interactions; - high number of parameters governing the lateral response of an infilled mesh of frame.
It follows that the models that have been proposed by different
researchers are strongly related to the kind of masonry infills that were
examined and to the experimental tests validating the models themselves. As regards this, Figure 1b shows how the section of the equivalent diagonal strut by different authors is differently related to the same
synthetic parameter lh’, which depends on the geometrical and
Fig. 1- a) Main published papers ; b) Ratios w/d proposed by different authors.
mechanical properties of the two sub-systems (frame and infill). In the
figure, w denotes the height of the section and d is the diagonal length
of the infilled mesh (Refs. 3-7). Further discordant results can be found
considering models including the post-elastic hysteretic behaviour.
Therefore, the main motivation of the present study lies in the nonavailability of an univocal approach, able to define an appropriate infill
model depending on the properties of the masonry utilized.
2.6 SCALE: Behavior and Strengthening of Stairs
In general the presence of a stair creates a discontinuity in a regular
reinforced concrete skeleton frame made of beams and columns; in fact,
from the geometrical point of view, a stair is composed by inclined elements (beams and slabs) and by short (squat) columns. These elements
contribute to increase the stiffness of the stair due to the elastic behaviour of inclined elements and of squat columns. For these reasons the
elements that constitutes the stair are often characterized by a high seismic demand: the squat columns are subjected to high shear force that
can lead to a premature brittle failure; the inclined beams, differently
from the horizontal beam, are defined by high variation in axial forces
that can modify the resistance and deformability of all these elements.
Although this is well known, no studies have been conducted by
researchers to evidence the role of stairs on the seismic capacity of
existing RC buildings; the identification of the weakest elements of the
structure and the failure type considering the presence of the stairs are
of particular interest. In this way, the knowledge of structural solutions
and design practice of stairs is an important step in order to define their
real geometric definitions and to understand their seismic performances.
2.7 NODI: Behavior and Strengthening of Beam-Column Joints
Observation of the damage caused by strong earthquakes on RC buildings designed to resist only to gravity loads showed that the main mechanisms that characterize structural collapses are, beyond the yielding of
primary elements such as column and beams, slippage of longitudinal
bars in columns and beams and joint failures (e.g. Braga et al. 2001).
Based on these observations and on the results of extensive experimenRESEARCH - Seismic behavior
tal campaigns, some provisions were inserted in the Italian technical
regulations imposing performance criteria for the design of new RC
structures placed in seismic zones. The capacity design approach provided by current Italian and European codes (NTC 2008, CEN 2004)
aims at preventing brittle failure mechanisms in beam, column and
joint members as well as at ensuring a weak beam-strong column global collapse, being more favourable in terms of overall ductility. For
existing RC structures, designed without anti-seismic criteria, there is
the problem of a reliable assessment of their seismic resistance also in
order to identify the more appropriate strengthening intervention systems. Improving knowledge on capacity models, particularly as for typical Italian building structures, is the principal thrust for the research
activity of task 7 (NODI) in the framework of DPC-ReLUIS 2005-2008
The workflow in terms of literature review, experimental testing and
numerical analysis performed by the RUs involved in the task is, then,
finalized to the analysis and validation of the provisions of Italian and
European codes in order to improve them and make them more adherent to the reality of the Italian existing RC building stock.
2.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
Modeling of reinforced concrete members is really a traditional topic of
structural engineering, but some aspects need further development
when seismic assessment of existing constructions is concerned. In fact,
well-established results for modern concrete structures do not cover a
large population of members built with obsolete materials and structural details like smooth bars. This is actually a relevant issue, since bond
between steel bars and the surrounding concrete is poor and anchoring
mechanical devices can play a relevant role in the development of plastic deformation and therefore of the drift capacity.
A number of models characterized by different models of complexity
can be found in the National and International technical literature (fib,
2003; Panagiotakos and Fardis, 2001; Park and Paulay, 1975) and provide an estimation of the rotation capacity at yielding and at failure of
columns member. However, they generally are able to well represent
response of r.c. members where deformed bars are used. Based on such
a background, the research on columns subjected to biaxial bending
and axial force has been conceived to cover the lack of knowledge at the
time of proposal. In fact, advances in seismic Codes and increasing
need of data for design purposes can be addressed among the primary
motivations of Task BIAX. On the other hand, since tools for the estimation of strength and deformation of bare cross sections were not so
consolidated, a specific focus on columns strengthened with FRP materials is certainly of applicative interest. This circumstances confirm the
RESEARCH - Seismic behavior
rational basis of the research and above all the actual usefulness of its
2.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
Precast structures passed through the check of weak and strong earthquakes and have been submitted to a wide specific experimental and
theoretical investigation performed in the main international research
centres. From these experiences some key aspects turned out to be
determinant for the good seismic behaviour of precast structures. These
key aspects are listed below:
- dry friction supports, not suitable to avoid the loss of bearing;
- diaphragm action, important to avoid joint distortions;
- lateral supports, necessary to avoid the overturning of beams;
- 2nd order effects, to be considered to avoid early collapses;
A positive condition of the existing buildings of concern is the possible
presence of a bridge-crane which required a structural design with relevant horizontal forces and a proportioning of the columns which could
be adequate also for seismic action even in the presumption of low ductility.
The regulation in force for the design of structures in seismic zones at
the time of construction is obviously a conditioning aspect which affects
the seismic capacity of existing buildings. Actions and rules for design
have been taken from that regulation which may result inadequate on
the base of the today knowledge. The problem concerns the seismic
zoning on one hand and the design criteria on the other.
In the following sections, the objectives pursued in each Task are
3.1 MND: Non-Destructive Methods for the Knowledge of Existing Structures
Research has been mainly focused on the evaluation of the role of the
main factors affecting the estimation of the in-situ concrete strength
through destructive and non-destructive tests, on the determination of
the design concrete strength, on the evaluation of the possible damage
on core specimens due to drilling and, finally, on the load bearing
capacity of structural members subjected to drilling before and after
restoration interventions. Other activities were mainly devoted to analyze the correlations between the various methods and the possible spatial variability of concrete properties throughout the surveyed members.
Regarding data collection, a large amount of experimental data from
destructive and non destructive in-situ investigations on real strucures
were collected and analysed, also through treatment of uncertain vari11
ables with different mathematical nature. In addition to the above topics, the seismic behavior assessment of buildings with structure composed of unidirectional RC frames was carried out by means of nondestructive in situ tests, with the objective of estimating their horizontal load-carrying and dissipative capacity. Finally, another important
objective was the evaluation of the dispersion of experimental results
from non-destructive measurements based on a critical review of data
reported in the literature.
3.2 FC: Calibration of Confidence Factors
This Task had two objectives. The first one was to propose a methodology for the calibration of the CF for materials strength, taking into
account the uncertainties characterizing existing building and the
effects on the reliability of the assessed structural performance. A
methodology was also sought for the evaluation of material strength by
destructive and non destructive in situ testing methods taking into
account the relevant reliability. The procedures were meant to be based
on the application of the Bayesian method. The proposed methodology
and the equation developed for FC have been validated on several simulated cases and on tests made on several buildings. The second objective was to develop a probabilistic methodology for seismic assessment
of existing buildings taking into account explicitly the uncertainties in
the material properties and the structural detailing parameters and
implementing the available test and inspection results. This methodology may be used for determination of confidence factors.
3.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
The main objectives are:
Validation of available modelling alternatives for RC buildings, mainly
lumped-plasticity and distributed plasticity models, both in commercial
and research software.
Validation of current methods of analysis for the seismic assessment of
existing RC buildings, with emphasis on nonlinear methods and their
applicability to plan-irregular buildings.
The above validations were carried out through the analysis of several
buildings selected by the different research units. Three buildings
(shown in Figure 2) were selected as common tested structures: one is
a doubly symmetric rectangular building, one is an L-shaped building,
and the third is a rectangular building with an internal court. These
buildings are representative of the structural buildings commonly found
in Italy. Several commercial and research programs were used for the
nonlinear analyses, including SAP2000, OpenSees, Midas, etc.
The final objective of this task is the compilation of a document that
contains: an introduction to nonlinear modeling of RC buildings and to
Fig. 2- Tested buildings.
the nonlinear methods of analysis of the European seismic codes: a
description of the main sources of nonlinearities in existing RC buildings: the application of different modeling techniques to the seismic
vulnerability assessment of the three building mentioned above. The
document is intended to be a primer for practicing engineers who want
to use nonlinear methods of analysis.
3.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
With reference to the first goal, technical literature and international
guidelines have been studied in order to define the classification of the
main geometrical characteristics of such kind of buildings and the study
of the response of mixed-type buildings – behaving as parallel systems
– subjected to horizontal forces.
With reference to the second goal, a series of numerical analyses based
on different and progressively refined modelling assumptions have been
performed in order to investigate the seismic action distribution between different technology elements, changing the size and then the
stiffness of the RC elements, but retaining the geometry of the building
and comparing the seismic behaviour of the mixed-type building with
the original masonry one. Pushover analysis have been performed by
using a lumped model to evidence critical zones and possible failures.
3.5 TAMP: Influence of Infills on Structural Response
A first experimental investigation was devoted to determining the
mechanical properties of three typical kinds of resisting elements, commonly used for infill masonry, and of the mortar utilized for their assemRESEARCH - Seismic behavior
bly. Then, several infill samples were subjected to compressive tests by
assuming orthogonal or parallel loading directions with respect to the
mortar layers. Further results were obtained under diagonal compressive loading, to determine shear modulus and resistance. At the end of
this phase, the experimental values of elastic moduli and resistances
were compared with the values that Italian M.D. ’87 provides by linking the mechanical properties of masonry elements to those of their
A second phase of the experimental research was devoted to acquiring
the response of square infilled meshes of RC frames subjected to a
cyclically varying lateral forces. Two 1:2 scaled samples were tested for
each of the three kinds of infill that had been mechanically characterized previously. The results of these tests have made it possible to calibrate the hysteretic model of pin-jointed diagonal strut proposed in
Cavaleri et al., 2005.
A further experimental investigation was carried out by means of shaking table tests on a 1:2 scaled 3D infilled RC frame, reproducing an
actual non-infilled building, previously subjected to pseudo-dynamic
tests at the ELSA-JRC-Ispra. These tests had the following objectives:
- to quantify the lateral stiffness and resistance contributions that infills
can provide; - to verify the influence of the infills on the crack distribution and the collapse mechanism. The same objectives were pursued
by nonlinear numerical analyses on multi-storey RC frames subjected
to natural seismic accelerograms. These analyses also showed the negative effects of non-uniform infill distribution along the height.
Another experimental campaign on infilled r.c. frames (1:1/2 scale), on
materials (concrete, steel, blocks and mortar), and on subassemblages
(small panels) has been performed. These tests had the objective of calibrating equivalent strut models, through comparison of experimental
results on bare and infilled frames, in order to evaluate the infill contribution as well as its uniaxial force-displacement relationship. The
constitutive models for masonry infills have been also calibrated in
order to predict the cyclic response of infilled frames.
3.6 SCALE: Behavior and Strengthening of Stairs
With reference to the first goal, several available manuals and books at
the time of construction have been studied in order to define the typology classification and the corresponding evolution of this classification
during the years with the increasing knowledge on the use of the materials and of computational machines. An analysis of the codes used
from 1909 to the 1980ies has been conducted in details with a critical
judgement based on the actual knowledge. Examples of stairs designed
for only gravitational loads have been studied with reference to different typologies.
With reference to the second point a series of numerical analyses based
RESEARCH - Seismic behavior
on different and progressively refined modelling assumptions and criteria has been performed in order to investigate the principal failure
modes. A critical study has been conducted on the different shear
strength formulations present in literature (Biskinis et al.2004; Sezen et
al. 2004; Zhu et al., 2007), in order to simulate potential shear failure
in squat columns, which can be easily found in most buildings. The
pushover analysis by using a lumped model has been performed to evidence critical zones and possible failures.
With reference to the third point, a test set-up has been defined in order
to investigate the experimental behaviour of a building sub-assemblages, including a stair substructure.
3.7 NODI: Behavior and Strengthening of Beam-Column Joints
A wide experimental campaign on beam-column joints representative
of typical members present in Italian existing buildings was planned,
designed and carried out. In particular, the research activities were
devoted to outline the influence of some parameters on the mechanical
behaviour and the failure mechanism of the joints, such as axial force,
amount of reinforcing steel and earthquake design level. Furthermore,
the research focused on the code expressions for the evaluation of the
ultimate rotation of RC elements in order to highlight possible discrepancies between the theoretical and experimental results. Another
research objective was the analytical modelling of beam-column joints
by using DIANA software to analyze the main parameters affecting their
seismic performance and, specifically, the analytical modelling of
experimental tests conducted on external joints. Other experimental
tests on beam-column joints relevant to existing buildings were performed as well, following an experimental program complementary to
the above-mentioned one, that is, in this case specimens reinforced with
smooth bars were tested and, in some cases, after the first test, joints
were retrofitted to evaluate the effectiveness of some retrofit systems.
Finally, supplemental activities followed two main branches: on one
hand, a series of either reinforced or unreinforced base joints were tested thus evaluating the performance of several strengthening systems,
and, on the other hand, a wide database of tests on beam-column joints
was built and analyzed.
3.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
The activities have been planned to cover four main objectives: (1)
review of technical literature with specific reference to available experimental data; (2) development of refined and simplified models for bare
and FRP reinforced members; (3) experimental activity on columns
subjected to cyclic actions; (4) drafting of a technical report summarizing the main applicative aspects of the work.
3.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
The study envisages a preliminary classification of the industrial prefabricated building typologies existing in Italy, from which the most frequent
characteristics of element-to-element connection types will emerge. This
first cataloguing phase is then followed by a purely experimental phase in
which some connections, identified as more vulnerable (e.g., friction connections), are subjected to a series of cyclic tests to simulate seismic conditions. Results and information obtained from the experimental tests will
serve as a basis to develop practical models for assessing the capacity of
such connection zones and to orient towards the definition of criteria and
techniques for strengthening interventions.
The main results obtained by each Task are summarized in the following sections.
3.10 MND: Non-Destructive Methods for the Knowledge of Existing Structures
The results obtained during the Project are mainly made up by the execution and analysis of experimental investigations either on in situ real
structures or on laboratory specimens, by the implementation of some
procedures to estimate the in situ concrete strength and by the uncertainty treatment of the structural characteristics of existing strucures.
A wide experimental program was carried out, comprising more than 20
RC beam and column members, several hundreds of non destructive
tests (NDTs) and more than 50 destructive tests (cores). Analysis of
results has shown a large scatter of the core concrete strength both in a
single member and among members extracted by the same story of a
building. Lower scatters have been detected for the NDT results with
the exception of the surface ultrasonic velocity (see Figure 3). As a
result of these findings, the role of some factors influencing the in situ
concrete properties has been carefully evaluated, and some criteria to
suitably select locations for sampling have been provided. A procedure
for the evaluation of the concrete strength based on the Sonreb method,
using both core and NDT measurements, has been set up and widely
validated, clearly showing its higher prediction capacity when compared to the relationships currently available in the technical literature.
It requires that the relationship between the in situ concrete strength
and the NDT measurements is experimentally derived for the specific
concrete under test.
As for the possible damage on core specimens due to drilling, the
results have shown that the strength reduction suffered by cores can be
significantly influenced by the original strength value of the in situ concrete. Consequently, adopting a constant coefficient to take into account
drilling damage, as suggested in the technical literature, can determine
incorrect results. On the contrary, it appears suitable adopting coefficient values, obtained during the research, which are inversely propor14
Fig. 3- Role of past applied loads on in-situ measurements: (a) qualitative bending moment due to vertical loads (a), and (b)
test results along the lower part of the extracted beam (rebound number S, direct velocity V, surface velocity Vs, core strength
tional to the core strength as provided by the compression test. Finally,
some important results regarding the effect of core drilling on the structural members, performing tests before and after a possible restoration.
Further, some factors influencing the relationship between the “local”
strength provided by core specimens and the in situ strength of the
structural member as a whole, have been highlighted.
Regarding the variability of concrete mechanical properties throughout
single structural members and among different sampling locations,
investigations based on a wide series of experimental data gathered
from surveys carried out on structures assessed for seismic vulnerability were carried out.
Main results obtained are briefly outlined below:
• no general trends have been recognized regarding the spatial variability of the key mechanical properties of concrete throughout column
members as a possible result of the combination of the effects of the
load pattern and the casting process;
• although carried out on members already cracked and damaged, the
results of sonic tests are affected by scatters smaller than those deriving by the compression tests on concrete samples; further, the ratio
between ultrasonic velocity derived by indirect and direct measures are
normally distributed around the average value of 0.75;
• rebound tests have confirmed the substantial impossibility of recognizing general trends in the spatial variability of the mechanical properties of concrete and led to values affected by scatters quite similar to
the destructive ones.
RESEARCH - Seismic behavior
Regarding the treatment of uncertainties in determining the characteristics of materials and more generally of existing building parameters, a
fuzzy-logic based approach for uncertainty treatment has been set up
and a computer code for its implementation has been developed.
In addition, the prediction capability of some formulations provided by
the current technical literature was verified based on experimental
investigations through non-destructive and destructive tests on existing
Based on a a critical review of available literature, a database was prepared that collects literature data on non-destructive tests on concrete
specimens for concrete grade assessment. Dispersion of experimental
results has been estimated and the influence of uncertainties coming
from the concrete grade estimation on seismic capacity of RC existing
buildings has been investigated.
Some results obtained during the project have been reported in papers
published on journals and in proceedings of Conferences (e.g. Masi and
Vona, 2008; Marano et al., 2008; Olivito et al., 2008).
3.11 FC: Calibration of Confidence Factors
A first aim of the task has been the evaluation of the confidence level
on structural safety of existing buildings given by seismic structural
assessment carried out according to the indications of the Italian OPCM
3431, 03-05-05, Annex 2.
Uncertainties in reliability structural analysis are due to material properties, structural details and condition of the structural elements. The
prior distribution of the considered uncertainties takes into account
their mechanical effects. The proposed probabilistic models are subsequently updated by in-situ information.
A parameter describing the structural performance is defined as the
demand/capacity ratio and its probability distribution is assessed by a
Monte Carlo simulation. Each realization corresponds to an application
of the capacity spectrum method and needs the execution of a structural linear static analysis. The Bayesian up-dating of the structural reliability is carried out by a Markov Chain Monte Carlo algorithm. The
structural performance prior probability distribution function is evaluated in two different cases: 1) taking into account the uncertainties
about material properties and structural details; 2) updating the structural assessment based on in-situ tests and inspections.
The updating process consists of two different levels: in the first,
destructive tests and relating errors are taken into account; in the second, non-destructive tests and relating errors are taken into account.
Structural seismic performance has been evaluated in three cases: a) by
using the mean material strength values (CF=1); b) by using the mean
material strength values scaled by the CF for a Normal knowledge
(CF=1.2); c) by using the mean material strength values scaled by the
RESEARCH - Seismic behavior
CF for a Limited knowledge (CF=1.35).
For each value of the CF the demand/capacity ratio has been evaluated
and positioned in the structural performance distribution, for the three
knowledge levels, with reference to the structure at hand. The CF value
considered as the exact one has been defined as the value corresponding to the demand/capacity ratio value with 5% exceedance probability. In this way, the CF for material strength is evaluated on the basis of
the probabilistic evaluation of structural performance taking into
account all the intrinsic and epistemic uncertainties, including uncertainties on the testing methods.
A simplified method is proposed too, based on a limited number of
Monte Carlo simulations, which is able to approximate the probability
distribution of the structural parameter. This can be a basis for the
development of simple procedure to use in for evaluation of structural
A second objective was to develop a procedure for evaluation of material strength and calibration of CF based on the application of Bayesian
method, to take into account the number and the reliability of the insitu tests carried out.
The Bayesian method allows to employ destructive and non-destructive
testing results to update a prior probability distribution function.
Destructive and non-destructive testing results are separately
employed, taking into account individual testing reliability (reliability
due to testing errors and errors in regression curve that provides the
material resistance as a function of the testing parameter). More than
one test method can be employed performing consecutive up-dating of
the probability distribution function.
The statistics reliability of the mean value is improved by applying the
confidence interval for the mean; a 95% lower confidence level is considered, which represents the value for the structural assessment.
In order to facilitate its evaluation, a simplified procedure is defined.
The material strength value for structural assessment can be obtained
scaling with an appropriate Confidence Factor a weighted mean of the
sampling mean values obtained by, both, different testing methods and
prior information:
mD = FC 艑 m̆'''inf,m̆f
m' +n x +n x
where xDM and xNDM are the sampling mean of the destructive and non
destructive tests, respectively; nDM and nNDM are the corresponding
sampling dimension.
Generally, if Mi is the i-th testing method adopted, the material strength
for the assessment is:
m' + n x
m = f i Mi Mi
where xMi is the sampling mean of the i-th testing method and nMi its
The CF can be expressed in an explicit form as a function of the
Bayesian coefficient of variation Vm for the median value of the material strength:
FC = a+cVwm
The parameter Vm, which estimates the reliability of the available information, is defined as:
i 2 Mi 2
Vm = m =
n x
i 2 Mi M2 i
where s2s,Mi and s2t,Mi are the sampling variance and the variance of the
regression curve of the i-th testing method, respectively.
The Eq. (4) has been calibrated for concrete strength by applying the
least squares method.
A Monte Carlo method has been used to simulate sampling with
destructive and non destructive testing and the resulting equation for
the CF is:
FC = 0.9 + Vm
The Eq. (6) is effective if samples have been extracted from homogeneous zones of the structure. If in the structure potential non homogeneous zones are identified, the t-Student test can be executed on the
mean values extracted from the two zones. If the t-student test identify
non homogeneous zones these must be separately evaluated considering two different median values for concrete strength with two CF.
A method has also been investigated for the evaluation of reliability of
the correlation function for the assessment of material strength by in
situ tests.
3.12 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
The main results of the project are as follows:
• Regarding the validation of the modified pushover procedure proposed by Fajfar (2002), it was found that it is conservative for plan-regular framed structures with respect to NTHA, while it is unreliable for
shear wall structures. For irregular frames, multi-modal procedures
have to be used in order to improve the accuracy of static non linear
analyses. Furthermore, predictions drawn from non linear static analy16
ses were often found un-conservative in terms of interstorey drifts or
chord rotations;
• A new pushover method that explicitly takes into account the torsional behaviour of asymmetric-plan buildings was defined. The effectiveness of the proposed procedure was evaluated by comparing the
seismic demand of selected case studies with that obtained through
both nonlinear dynamic analyses and other pushover methods;
• The nonlinear analyses on the regular and irregular buildings have
shown the importance of damping in nonlinear dynamic analysis. More
specifically, as the hysteretic model improves, the damping should be
decreased. Furthermore, viscous damping is hard to assign and little or
no indications are given in the published literature to guide the user. A
value of 2-3% damping for the elastic modes appear reasonable, but no
final indications were found;
• In NTHA, for both natural and generated accelerograms, the application of the seismic input in the principal directions of the structure may
underestimate the demand, the structural demand varies considerably
as the seismic input direction changes, more so for natural accelerograms. However, as for the number of input ground motions to use in
nonlinear dynamic analyses, enhancements to the EC8 requirements
were proposed;
• For irregular buildings, pushovers applied in different directions indicated demands and capacities that depend on the direction considered.
Partial results on the L-shaped building are shown in Figure 4.
• Shear collapse in existing buildings not designed according to the
capacity design rules, is often dominant. If only the flexural capacity is
modelled, most frames are verified both at the Ultimate and at the Near
Collapse Limit State. However, additional checks show that the shear
capacity has already been reached on several elements prior to reaching the design accelerations. This indicates the need to use models that
consider shear failure too (this is very rare in the available software).
Furthermore, early shear failures point to possible retrofitting actions,
Fig. 4- Pushover analyses on L-building: Influence of seismic input angle on the demand: displacements (left) and rotation
ratios (right).
RESEARCH - Seismic behavior
such as shear strengthening of beams and columns;
• Synthetic expressions were developed that express the plan-regularity of a building with respect to stiffness and strength distributions.
Also, simple corrective coefficients were developed to compute the
structural demand increase as the structural eccentricity increases. In
alternative, a simplified procedure was developed that can help optimize the structural design by producing a plan-regular building;
• Severe convergence problems were encountered in commercial codes
that use lumped plasticity models. These problems were related to the
software limitations rather than the modelling selection. As expected,
fibre section models provide a better, more physically-based prediction
of the section response. It is however important to extend the model to
include shear failure, in order to avoid post-processing of the results;
• Different Limit State definitions for performance assessment were
considered, such as interstorey drift, plastic rotation and chord rotation.
It was shown that different measures lead to different, sometimes very
different, capacity predictions. The limits usually assumed for interstorey drift result in larger chord and plastic rotation than limits proposed by the European and American Codes. Furthermore, the procedure provided by Eurocode 8 to compute the chord rotation capacity
yields predictions that are very different from those obtained analytically, that is integrating the member curvature throughout the plastic
hinge length;
• A simplified method for deriving fragility curves and evaluating the
probability of failure was proposed. The method is based on an incremental application of the so called N2-Method with natural response
spectra, whose irregularity covers the record-to-record variability of the
structural response without the need for performing non-linear dynamic analysis.
3.13 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
The main activities of the first goal have involved both the literature
review of mixed-type buildings and the modelling aspects related to
the RC substructure of such structures. A report was produced
(Decanini et al., 2006) including the main results obtained in the
The work involved the study of 160 papers, as indicated in Table 1.
Confined masonry is listed separately to underline the difference in
the amount of studies between those constructions and structures in
which elements of different technologies (masonry/reinforced concrete) are not bonded. The latter case includes constructions in which
masonry and reinforced concrete frames are present at the same level
and construction in which reinforced concrete frames are placed in
the upper floors only.
RESEARCH - Seismic behavior
Table 1. Literature review on mixed-type buildings
Damages after earthquake1
Mixed-type building2
Analytical studies
Confined masonry3
Analytical studies4
Reports and other papers enclosing field observations after earthquakes.
Building in which elements of reinforced concrete and masonry are not bonded.
Building in which bearing masonry walls are confined by reinforced concrete elements.
4 Analytical studies, including also proposal for design and for the modelling of confined
masonry structures.
The main conclusions inferred by the study can be summarized as follows. In many codes the confined masonry structures are taken into
consideration, even though usually just empirical dimensional rules are
given, while methods for assessment and design are lacking and, when
present, they differ substantially from a code to another. Concerning the
mixed structures (unconfined masonry), only a small number of codes
gives some indications, which are usually about the behaviour factor
and/or the distribution of horizontal forces among different structural
elements. A similar trend was found also when gathering papers on field
observations, experimental tests and analytical studies: a lot of data
were found concerning confined masonry structures while just a few
were gained about other kinds of mixed structures. In general it is possible to state that confined masonry can withstand to seismic actions,
given that materials are of good quality and good constructions rules are
followed. On the contrary, some deficiency, such as small amount of
transverse steel in columns, high thickness of mortar joint, lack of RC
element near the openings, can generate the bed behaviour of this kind
of structures. Concerning the mixed structures (other than confined
masonry), the main problems which arise in the assessment and design
are related to the distribution in plan of the horizontal actions, to the
determination of the behaviour factor and to the study of the connections between masonry and reinforced concrete elements. In literature,
indications about these points are few and further studies are needed.
With reference to the second goal, the main problems concerning the
non-linear analyses on RC-masonry mixed buildings have been highlighted. Particular attention has been devoted to the seismic action distribution between different technology elements. The results of the
analysis carried out on 3D RC-masonry mixed building (see Figure 5)
with external walls and internal frames, underline the fundamental
masonry role to withstand horizontal action, while a significant translation increase could be offered by introducing RC frames. This research
has shown the growing capacity offered by mixed building to bear the
seismic action by increasing the RC elements stiffness. The increasing
Fig. 5- Analyzed models (Nardone et al., 2008).
RC elements stiffness is significant for the seismic action distribution
both in linear and non-linear field. From the analyses it has emerged
that, by increasing the stiffness of internal RC frames, the maximum
sustainable seismic action of mixed building increases, while the rate
of seismic action supported by masonry walls decreases (Nardone et al.,
2008). The important slab role in sharing the seismic action between
the vertical resistant elements and the importance of slab to perform
this function in order to avoid undesirable behaviour of the building has
been underlined.
In addition to that, concerning the first model, the analyses highlighted
that the presence of the RC frame is generally detrimental for the
masonry wall. In fact, the resisting total base shear of the mixed system
is smaller than that of the masonry wall alone when the contribution of
the frame wall is smaller than about 15-20% of the total base shear.
Moreover, when the top connection is non effective, hinges develop at
the top and the bottom of RC columns. Therefore, it seems appropriate
to design the RC elements considering seismic action deriving from the
pertinent vertical loads. While, if the connection between masonry and
frames are effective, it would be expedient to assign the whole of the
horizontal loads to the masonry (Decanini et al., 2008).
In general, the analyses performed on the 3D model highlighted the difficulties in modelling the masonry and the connections between different elements (beam-masonry, floor-masonry) and indicated the strong
influence on the global structural behaviour of the connections effectiveness. The parametric analyses showed the importance of the masonry tensile strength among other mechanical parameters on the global
response and highlighted that the RC elements remain elastic and give
a negligible contribution to the overall performance. Finally, the comparison between the mixed structures results and those obtained for the
masonry structures confirmed how the practice of replacing masonry
walls with RC frames, in the interior of old masonry buildings, can have
negative consequences on the vulnerability of the building themselves.
3.14 TAMP: Influence of Infills on Structural Response
The resisting elements that were chosen to construct commonly used
masonry infills were: calcarenite ashlars, hollow clay brick and hollow
lightweight concrete blocks. The results of the compressive tests on the
related masonry samples lead to the following main conclusions: i) calcarenite masonry exhibits middle compressive and shear strengths, and
ductile behaviour up to the collapse for each of the three considered
loading directions; it behaves like an orthotropic material in which the
ratio between the two elastic moduli depends on the mortar properties;
ii) clay brick masonry is about twice resisting in compression, but ductile behaviour was observed only under diagonal loading, when the
shear behaviour of the mortar joints is involved; iii) lightweight concrete
masonry has very low resistance capacities along all the loading directions; the results obtained from this kind of masonry were rather scattered, because vertical mortar joints are not provided for, due to the vertical profile of the resisting elements. The experimental values of elastic moduli and resistances showed that the values deduced by the M.D.
’87 provisions are not always reliable. In particular, the shear strength
values are strongly underestimated by this code for all masonry tested.
The cyclic tests on the infilled meshes of RC frames have made it possible to verify that the cross-section of an equivalent strut can be determined by using the procedure and parameters shown in Amato et al.,
2008, where a not negligible role is exerted by the transverse strain
ratio in the diagonal direction and the compression level transmitted to
the columns after the masonry infills have been constructed. The lateral resistance of the infill can be deduced from the masonry shear
strength; this can be translated into a compressive strength to be conferred to the equivalent strut, by means of a suitable criterion taking
into account mainly the disconnection arising between frame and infill.
Finally, these tests showed that the calibration of the adopted model
leads to sufficiently approximate results (Figure 6).
The shaking table tests on the 3D scaled building (Figure 7) have been
carried out by assuming a natural input accelerogram (Herceg-Novi,
Fig. 6- Cyclic tests on infilled RC frames: a) test set-up; b) experimental results; d) validation of model.
RESEARCH - Seismic behavior
Fig. 7- Results of shaking table test (PGA = 0.3 g): a) 3D building; b) crack distributions.
Montenegro 1979), scaled to three levels of PGA. These tests showed
that: - the fundamental period of vibration of the structure is influenced
by the presence of the infills (PGA = 0.04 g); - the crack distribution
depends on the location of the infills with opening (PGA = 0.3 g); - the
infilled structure bears a PGA value (0.54 g) that proves to be about
three times the value that had been deduced by pseudo-dynamic tests
on the corresponding non-infilled structure.
With regard to the numerical investigation, two series of four-storey and
twelve-storey RC frames have been considered. Frames of the first
series were designed to bear only gravitational loads; the ones of the
second series were designed according to the EC8 provisions. The
results lead to the following remarks: - the presence of infills implies
greater quantity of input energy for the structure; nevertheless, the
infills can dissipate a lot of this energy so that the total balance is
favourable to the resisting elements of frame with respect to the case of
bare structure; - in the frames of the first series the infills can exert a
decisive role towards a seismic event; - in the frame of the second series
a non-uniform distribution of infill in elevation would be considered in
defining the structural regularity.
Finally, the studies regarding the calibration of a three-strut model
arrived at the conclusion that this approach is useful for the comprehension of the local behaviour of columns in case of degradation due to
infill-induced shear effects. For what concerns response in terms of displacement, single strut models seem to give satisfactory results. Strut
parameters can be directly obtained from the mechanical characteristics of infill components, i.e., mortar and blocks. Moreover, a design criterion for a protecting dissipative bracing systems has been developed.
The proposed approach allows to define the required minimum dimenRESEARCH - Seismic behavior
sions giving the desired interstory drift.
The results of this Task, synthetically presented here, are discussed in
detail in the final Report of Task TAMP. The following remarks can be
made, concerning their use and possible improvement of provisions: the proposed model of equivalent diagonal strut could be utilized for
analyses of existing structures designed to bear only gravitational loads;
- the expressions proposed by M.D. ’87 would be revised, also including values to be assigned to the transverse strain ratios; - infills would
be considered for evaluation of the period of vibration of the structure
and definition of structural regularity.
3.15 SCALE: Behavior and Strengthening of Stairs
According to the literature, the existing stairs can be classified into two
main categories depending on the static behaviour of the stair steps: (i)
stairs with steps performing as cantilever beam, stair type A (see Figure
8a), (ii) stairs with simply supported steps stair, type B (see Figure 8b).
Generally the stair type B are used worldwide, in Europe and USA,
while the stair type A, with inclined beams, are much more adopted in
Europe (Tecnica y Pratica del Hormigon Armado, 1989, Reynolds and
Steedman, 2002, Guerrin and Lavaur, 1971). According to the USA’s
manuals a great scatter in stair structural solutions can be found (Berry,
According to the manual design criteria (Marrullier, 1910; Rosci, 1939;
Santarella, 1953, 1957; Pagano, 1963; Migliacci, 1977) stair type A
could be designed considering only gravity loads, any seismic actions
could not be taken into consideration. The permanent and live loads on
the steps generate on the beam element (bs1-bs2-bs3) a torsional
moment T and a distributed load producing on the beam shear force V
and bending moment M. Each flight step is designed modelling it as a
cantilever beam subjected to a distributed load.
As it is explained in several dated manuals (Santarella, 1953, 1957;
Pagano, 1963), the design bending moment into the beam (bs1-bs2bs3) is evaluated on the basis of different static schemes corresponding
to different constraints at the extreme ends of the beam. In particular,
two extreme constraint conditions are suggested: full constraint and
simply supported. The torsional moment is considered of relevant
importance, it leads to add transversal reinforcement (stirrups) along
the length of the beam (bs1-bs2-bs3). The adopted values of the torsional moment depend on the hypothesis upon the flexural stiffness of
the inter-storey slab: flexible and rigid diaphragm.
About stair type B, manuals indicate two limit structural schemes: (i)
an horizontal beam full constraint at the end, (ii) an horizontal beam
simply supported at the extreme ends. Normally bending moment and
shear are the internal forces taken into consideration. In the manuals of
the construction time the only severe prescription is regarding the
Fig. 8- Stair typologies: (a) stairs type A with steps as cantilever beams, (b) stairs type B with simply supported steps.
Fig. 9. Results of the push-over analysis in terms of base shear versus roof displacement (Cosenza et al., 2008).
design of the steel bars: a reinforcing bar should not bend to form angles
that favour pull-out of the concrete cover (Pagano, 1963).
The Italian stair design practice during the period 1954-1980 has been
analyzed in order to identify the most common typologies and the effective adopted design criteria. As already remarked, according to the
dated technical manuals and codes stairs could be designed for only
gravity load. The predominant stair type in the studied building sample
is type A; the flight steps are cantilever elements constraint to inclined
beams having one point of discontinuity in the 53% of the cases, two
points of discontinuity in the 37% of the cases and is directly connected to the column without any discontinuity in the 5% of the cases. The
stair type B is present in the sample with incidence of 3%. The design
practice of the most common type of stairs, composed by flight steps
constrained into a beam, is herein studied. The static design scheme
has a great scatter; beams were designed considering a maximum
moment M+=qL2/a in the midspan with a=12 (30% cases) or a=8
(30% cases), while the minimum moment at the extremes of the beam
is obtained with a=12 in most cases (76% cases).
Regarding the influence of stairs on the seismic capacity, this preliminary study on the structural typologies of the building sample has evidenced the following problems related to the presence of stairs: distribution of seismic forces (not considered in the design), different modelling design of stair structure, material strengths, element detailing.
The structural typology of stairs generally introduces discontinuities
into the typical regular reinforced concrete skeleton, composed by
beams and columns; in fact, the sub-structure “stair” is an assemblage
of inclined elements as slabs or beams. All these elements contribute to
increase the stiffness of the stair due to the elastic behaviour of inclined
elements and of squat columns. For these reasons the elements that
constitute the stair are often characterized by a high seismic demand:
the squat columns are subjected to high shear demand that can lead to
a premature brittle failure; the inclined beams, differently from the horizontal beam, are defined by high variation in axial forces that can modify the resistance and deformability of all these elements.
All these aspects are discussed with a series of analysis on a RC build-
ing representative of the studied sample; non linear static analyses (static push over analysis) finalized to the evaluation of the role of the stairs,
of their elements and modelling is performed.
The building without any stair is defined as reference. Two models have
been considered to study stair type A with inclined beams and stair type
B having reinforced concrete slab. For each structure, different modelling have been adopted to evidence the influence on the global response
of: biaxial bending modelling in the beams of the substructure “stair”;
bending moment-axial force (M-N) interaction into the inclined elements (beam and slabs); bending moment-shear (M-V) interaction into
the inclined elements and columns.
In general, the presence of stair brings to an increase of lateral strength
and to a reduction in displacement capacity with respect to the building without stair (Cosenza et al., 2007a). On the contrary, the results
have confirmed the need to utilize biaxial bending modelling and to
account for the interaction of the different internal forces (Cosenza et
al., 2008) as: bending moment-axial force interaction that characterizes
the inclined elements, and the bending moment-shear interaction that
governs the behaviour of squat columns. Shear failure becomes predominant in the squat columns and in the reinforced concrete slabs and
precedes the conventional ductile failure (see Figure 9).
3.16 NODI: Behavior and Strengthening of Beam-Column Joints
Main results obtained in this Task are made up by the execution and
analysis of the experimental program on beam-column joints without
strengthening, as well as by the execution and analysis of some tests on
strengthened and unstrengthened beam-column joints and on base
joints. Numerical-experimental comparisons have been performed, as
well, regarding some highly representative test campaigns available in
the literature and relevant to some of the tests carried out. A careful literature review was carried out relevant to: (i) experimental programs
carried out by other researchers, (ii) joint capacity models, and (iii) code
provisions on beam-column joints.
A wide experimental program was carried out on full scale external
beam-column joints relevant to typical existing RC buildings having
RESEARCH - Seismic behavior
Fig. 10- Test results on a Z2 joint (design for medium seismic zone, low axial load): a) damage pattern at drift=7%, and b)
force-drift relationship.
different Earthquake Resistant Design (ERD) level. Quasi-static tests
have been performed with 3 loading cycles for each drift value gradually increased from 0.25% up to the total failure of the joint. Results
showed yielding force equal to about 20 kN for the NE joints (gravity
loads only designed), and about 40 kN for both Z2 joints (designed for
seismic zone 2, medium seismicity) and Z4 joints (designed for seismic
zone 4, very low seismicity), as a consequence of the minimum requirements on reinforcement amount prescribed by the Italian code.
Observed failure mechanism in all the joints showed a wide and heavy
cracking in the beam, due to the small amount of the longitudinal reinforcing bars as the beams were not loaded by the floor slabs in the considered building model. Joint failure was generally caused by the tensile failure of the reinforcing bars in the beam. A different as well interesting behaviour was displayed by some tests on Z2 joints (Masi and
Santarsiero, 2008), where a wide cracking also in the joint panel and a
softening mechanical behaviour (Figure 10) were observed, due to the
reduced amount of the applied axial force (=0.15). Drift value (coincident with chord rotation value) at failure in the NE joints is about 3.0%,
while in more ductile Z2 and Z4 joints, values equal to about 4.5% have
been detected.
As for the contribution of joint panel own deformations to the total drift
of the sub-assemblage, experimental results showed that it is rather low,
being always lower than 10% even in the heavily damaged specimens.
Further, interesting results have been found by comparing the strength
of the joint panel provided by the tests and that one obtained applying
the European (CEN, 2004) and the Italian Code (NTC, 2008) expressions to evaluate the capacity of beam-column joints. Results show a
good estimation ability of code expressions that have been able to predict which of the specimen was subjected to diagonal cracking of the
joint panel. Regarding the ductile capacity, the difference between NE
and seismic specimens was lower. In all the tests with high axial load a
failure mechanism with extensive cracking of the beam and evident
deterioration of the concrete at the beam-column interface has been
RESEARCH - Seismic behavior
A parallel experimental program was devoted to the test (under imposed
increasing cyclic displacements applied to the beam end) of 4 external
reduced scale beam-column joints provided with smooth bars, later
retrofitted after a first series of tests. For 2 of them only the anchorage
of the beam bars had been restored by welding threaded bars to the
ends of the longitudinal beam bars and bolting them to steel plates
placed on the column external surface. For the other 2 joints, besides
restoring the anchorage, also vertical and horizontal carbon fiber fabrics
had been applied on the column, below and above the joint panel.
Moreover, other 2 real scale beam-column joints were built and tested
under cyclic loads, either with or without retrofitting, to evaluate the
increase of joint performance. As for the first 4 joints, the restoration of
the anchorage of the beam bars proved to be very efficient, since it provided increases in strength up to 300% with respect to not retrofitted
joints. The additional carbon fiber reinforcements did not provide
noticeable increases in strength, because the cracking of the joint,
which would have required the carbon fiber contribution, did not occur.
Regarding the other 2 joints, it was observed that the not-retrofitted one
attained low strength values due to the lack of steel reinforcement for
negative moment in the beam. As regards the joint retrofitted before
testing, the carbon fiber fabrics applied on the beam significantly
increased both strength and displacement ductility.
Additional research activities were mainly focused on experimental
tests on full scale RC columns (base joints not-strengthened and
strengthened) tested under constant axial load and monotonic or cyclic
flexure. The strengthening systems consisted in confinement by partially wrapping unidirectional carbon or glass layers around the element
at the base. On some specimens, in addition to confinement, steel
angles (in some cases, anchored at the foundation) were placed in correspondence of the member corners. The specimens were designed
according to the Italian codes in effect during the ’60s and ‘70s with the
aim of reproducing typical dimensions, rebar amount and details common at that time. Main results achieved during the Project are as follows: (i) regardless of the axial load value, the FRP confinement produces significant increases in terms of ductility, especially if a GFRP
(glass) jacket is used; (ii) the arrangement of the longitudinal steel
angles unconnected to the foundation leads to higher ductility levels
than those measured for members strengthened by only FRP systems;
this ductility gain is lower for columns tested under n=0.40, even if in
these cases the unconnected angles also provide an improvement of the
flexural strength; (iii) when the longitudinal angles are anchored to the
foundation the flexural strength of the RC columns significantly
increases, but a reduction of the available ductility is observed.
A numerical-type activity focused on the analysis of the performance of
RC beam-column joints through numerical simulations by using the
F.E. software DIANA, validated by means of experimental test campaigns available in literature (e.g. by the Shiohara working group) as
well as some of the tests performed. In particular, many non linear
analyses have been performed on typical existing external beam-column joints as they can be found in real buildings built in the past. The
definition of typical deficiencies found in real beam-column joints and
the analysis of the main parameters governing the structural behaviour
of such joints allowed to highlight some effective strengthening solutions, especially for external joints. Further, such work allowed to validate some theoretical models able to predict the behaviour of both
external and internal joints, as well as to validate the expressions proposed by some codes to predict the failure of the joint panel.
Some results obtained during the project have been reported in papers
published on journals and in proceedings of Conferences (e.g. Masi et
al., 2008).
3.17 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
The presentation of the results follows the research outline that has
been above discussed. In particular, methodological and numerical
contributions are presented separately from the main experimental
findings. This choice is actually related to the nature of the results and
their impact on applicative aspects of structural seismic design, namely codes and design and assessment practice.
In particular, the detailed review of technical literature that has carried
out during the early stages of the research pointed out the relevance of the
type of reinforcement used for construction of existing buildings. In fact,
it has been demonstrated that experimental response of r.c. members can
be affected by type of reinforcement, smooth or deformed, and by bond
interaction between steel and concrete especially in post-yielding phase.
As a consequence, a concrete effort has been devoted to perform a comparative analysis of inelastic performances of members depending on
type of reinforcement; this means that both numerical and experimental
activities have been calibrated in order to cover at local -strength and
ductility of cross sections, bond under static and cyclic loads of rebarsand at global – stiffness and rotation capacity of members-.
Results of both theoretical and experimental activities have been published in the context of National and International conferences and
meetings. Specific attention has been paid also to continuous education
of young engineers and technical updating for practitioners. Diffusion
of software packages (free download of executables and tutorials) and
research results has been supplied on Reluis website.
The first set of results refer to the development of methods for numerical analysis of cross sections subjected to generalised bending and axial
force. Different approaches have been proposed by the different teams
involved in the Task, comparative analyses have been carried out and
multiple applicative perspectives have been covered.
A specific software has been developed and made available at the project website (www.reluis.it) for download. It is based on a Fortran language procedure and takes advantage of a user-friendly Visual Basic
interface and multiple language platforms. The computational engine of
the software has been also used for the development of an extended
parametric analysis aimed at the evaluation of the influence of the biaxial actions not only on the ultimate strength but especially on the cross
section ultimate curvature. In particular, such influence was studied on
square RC cross-section characterized by different values of axial load
and geometrical percentage of reinforcement. The study has been
developed in order to define simplified analytical formulation to easily
predict the ultimate curvature reduction in the case of biaxial bending
and axial load with respect to the case of uniaxial bending. Figure 11
gives a view of the program interface (top right), an example of typical
results in terms of generalised relationship between the reduction of
ultimate curvature compared with the reference uniaxial value
(=u,biax/u,uni), normalised axial load and the inclination of the
Fig. 11- Effect of biaxial bending on cross section local deformation (Di Ludovico et al. 2008a, 2008b).
RESEARCH - Seismic behavior
Fig. 12- Comparison between exact (fibre method) and approximate approaches for bare (left) and FRP reinforced r.c.
members (right) (Monti and Alessandri, 2008).
stress plane angle (top and bottom left) and finally a comparison
between simplified and refined results (bottom right).
An alternative method has been proposed and implemented. It arrives
at defining closed-form equations for performing assessment of existing
RC columns with two-way steel reinforcement, under combined biaxial
bending and axial load. Starting from the load contour method, an efficient procedure for estimating the strength/deformation section capacity has been developed. In addition, simple closed-form equations for
computing section uniaxial resisting moments and ultimate/yielding
curvature has been defined.
The method has been also extended to cross sections reinforced with
FRP, resulting in a very effective guide for fast implementation of
results in more general software packages and direct application by
structural engineers using easy to manage electronic sheets.
Figure 12 shows an example of results that can be obtained according
to this design tool, both for bare and FRP-strengthened members and
its ability to give reliable results with a reduced computational effort.
Another relevant achievement is related to the response of the whole
member under axial force and biaxial bending. In particular, the attention has been focused on the development of the actual stiffness of the
member depending on the stress level and influence of biaxial bending.
A fibre model of the member has been implemented in order to provide
recommendations for characterization of equivalent stiffness to be used
in elastic analysis. The study carried on has been limited, at the
moment, to the definition of the problems which must be taken into
account while passing from a cross section to the whole member, Bosco
et al. 2008. In this context, it is worth noting the contribution of UNICH
that carried out nonlinear analyses of reference regular and irregular
reinforced concrete buildings and on problems related to the selection
of input ground motion, Canducci et al., 2008.
In compliance with the main issues derived from the theoretical analysis of r.c. members and cross sections, experimental activity has been
carried out at different scales.
RESEARCH - Seismic behavior
Fig. 13- Fibre model of the cross section (left), force-displacement curve for a column (center), normalised equivalent stiffness
(right) (Verderame et al., 2008a, 2008b).
At local scale, a number of bond tests on smooth rebars were devoted to
the definition of a constitutive stress-slip law to be used in numerical
Figure 13 reports sample data concerning specimens (left), typical
cyclic experimental data (center) and finally the idealized constitutive
law calibrated against the test (right).
At large scale, eight experimental tests on r.c. square or rectangular full
scale columns under constant axial load and uniaxial bending were performed. Monotonic or cyclic action were applied on specimens. Details
about the experimental program are reported in Table 2 and Figure 14.
Each test was performed under displacement control.
Table 2. Summary of experimental tests on r.c. columns
UNIAXIAL TESTS - Cross Section BxH (cmxcm)
Longitudinal Reinforcement
Steel Rebars type
Normalized Axial Load, 0.2
Type of action
Number of tests
Primary experimental outcomes clearly indicate that the contribution of
the base rotation on the global deformation mechanism is noticeably
different in case of columns reinforced by plain or deformed rebars,
however, the overall member global deformation and energy dissipation
capacity is not strongly affected by the bond performances of the internal rebars. The global deformation capacity in the case of plain rebars,
is mainly due to a localized source of deformability at the column foun23
Fig. 14- Experimental tests on columns.
dation interface (fixed end rotation), while a more diffused crack pattern
along the column end was observed in columns reinforced by using
deformed steel rebars. The significant influence of P- effect on the
global behavior of specimens has also clearly emerged by the experimental tests; if such effect is neglected, the ultimate rotation recorded
on columns reinforced by deformed steel rebars is clearly less than that
observed in columns reinforced by using plain rebars. However, due to
P- effect the ultimate rotations related to the two different columns
typologies is very close. Such result can be explained by considering
two main aspects: the higher strength of members reinforced with
deformed rebars and the higher slope of the softening branch of the
shear-drift curves, if P- effect is considered. A calibration of a numerical model able to take account of specific aspects related to bond of
smooth rebars and anchoring end details has been also developed
(Verderame et al., 2008c).
3.18 PREFAB: Behavior and Strengthening of Prefabricated Industrial
The first activity consisted of a wide survey of the existing buildings
produced from the ‘50s up to today and of the rendering in a reasoned
way of the investigation results. This activity enjoyed the support of the
National Association of prefabrication industries Assobeton which
involved ten member companies to provide the design documentation
of a number of constructions built in different times. So about 150 project documentations have been collected covering some decades of production. Since information about far times were lacking, this survey has
been integrated with the historical memory of some experts, exploiting
their knowledge together with the old bibliography of specific journals.
From the examination of the project documentations a synthetic
description for any building has been recorded in a standard format,
summarising its features in a specific form of easy reference. This work
led to the printing of the booklet “Precast structures: list of projects of
existing buildings” which provides a good evidence of some decades of
precast production.
A complete catalogue of the different types of precast structures has
Fig. 15- FrontPage cover of two produced catalogues on precast concrete buildings.
been drafted and published in a specific booklet (Figure 15). Every type
has a short description with sketches and indications about the years of
production, the regions of destination and the relative diffusion with
respect to the global precast production. Some notes are added with
indication of possible behaviour deficiencies for seismic destinations.
Being a key point of precast construction, special attention has been
addressed to the seismic behaviour of connections. Some tests have
been performed to quantify their seismic capacity following a standard
Five principal categories of connections have been considered:
- floor-to-floor connections between adjacent floor or roof elements;
- floor-to beam connections between floor or roof elements and the supporting beam;
- beam-to-column connections between the beam and the column;
- column-to-foundation connections which provide the base support to
the columns;
- cladding-to-structure connections for the support of the wall panels.
Two level of tests have been performed:
- particular tests: referred to the qualification of single connectors
inserted between two overdimensioned blocks and subjected to the
principal action expected in the structural system (Figure 16, left);
Fig. 16- Details of testing of structural connection.
RESEARCH - Seismic behavior
- local tests: referred to the qualification of the connection included
between two significant portions of the elements, representing the structural arrangement and subjected to the relevant components of the
action (Figure 16, right);
Previously a standard protocol for testing has been drafted. It defines
the six parameters:
- strength: maximum value of the force which can be transferred
between the parts;
- ductility: ultimate plastic deformation compared to the yielding limit;
- dissipation: specific energy dissipated through the load cycles;
- deformation: ultimate deformation at failure limit;
- decay: strength loss through the load cycles compared to force level;
- damage: residual deformation at unloading compared to the maximum
to be measured both by:
- push over tests following a monotonic increase of displacement:
- cyclic tests following an alternate history of displacements (Figure 17).
A number of tests on roof-to-beam connections (8 push-over and 11
cyclic) have been performed. A complete report is available with the
results of the experimentation. Three types of steel connectors have
Fig. 17- Specific energy and cyclic behaviour of the connection.
been examined (Felicetti et al., 2008).
In parallel, a testing frame has been set up for push over and cyclic tests
on beam-to-column connections. The common type with a couple of
bars protruding from the column and passing through the holes of the
beams is examined both in longitudinal and transverse direction.
The experimental test results in terms of force displacement allowed to
characterise the connection roof element to beam, as global ductility,
ultimate strength and dissipative capacity. In Figure 18, force-displacement curves are illustrated (monotonic test on rigid blocks) comparing the traditional existing connection with a modified solution of the
connection. While a brittle behaviour is expected for the traditional
existing connection (concrete crash of the beam edges), a ductile behaviour is then obtained if the steel plate in the connection is opportunely
reduced. Almost similar conclusions can be obtained for quasi-static
cyclic tests as reported in the research reports and papers.
Some other tests have been performed on dry bearing in order to measure the friction factor of neoprene pads over the concrete surface and
their deformation parameters. Tilting tests and on inclined plane have
been made with different levels of normal loads (Magliulo et al., 2008).
The comparison between tests results and friction coefficient values
provided by PCI Handbook (1999), CNR 10018 (1999), and UNI-EN
1337:3 (2005), is shown in Figs. 19 and 20. In Figure 19 the neoprene
compressive stress (s) is reported on the horizontal axis, while the shear
one () on vertical axis; in Figure 20 on this axis the friction coefficient
is presented. It is evident that PCI Handbook and CNR 10018 curves
well approximate the experimental data linear regression curve; this
does not happen in the case of UNI-EN 1337:3 curve. However CNR
10018 provides a bit larger friction strength with respect to the experimental results, while PCI Handbook provides larger friction strength
only for compressive stress lower than 3 N/mm2. Furthermore, the tests
results confirm the light increment of friction strength, which corre-
Fig. 18- Push-over test: comparison of the two different solution (traditional vs. slightly modified).
RESEARCH - Seismic behavior
sponds to a light decrement of the friction coefficient, as the compressive stress increases.
On the base of experimental results, the following relationships for neoprene-concrete friction coefficient are proposed:
m= 0.49
if s 0.15N/mm2
if 0.15 < s 5N/mm2
m = 0.1+ s
where s is the compressive stress and =0.055 N/mm2; s=5 N/mm2
is the neoprene maximum compressive strength according to CNR
10018. These formulations along with the linear regression curve of
tests mean results are plotted in Figure 21: the two curves are almost
Fig. 19- Comparison between compressive-shear stress curves provided by PCI Handbook, CNR 10018 and UNI-EN 1337:3
and tests regression
4.1 MND: Non-Destructive Methods for the Knowledge of Existing
The research activities scheduled within the Task have been to a great
extent carried out without significant delays or anticipations, and the
main objectives have been achieved.
4.2 FC: Calibration of Confidence Factors
The results obtained agree with those expected, concerning the development of a Bayesian procedure for material strength evaluation and
the calibration of confidence factors.
4.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
The results obtained by the Task are in line with the objectives originally established. Several papers by the different research units have
been published or are in press in international journals or conferences.
The only problems encountered by some of the units originated from
convergence issues in nonlinear codes. This is a well-known problem,
but some software failed to converge on a regular basis, thus delaying
advances in the research. However, overall, the task followed the schedule of work originally outlined.
4.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
The Task has pursued the proposed objectives. In particular, aspects of
modelling the behaviour of mix-type buildings through non linear analysis have been investigated. These analyses have highlighted the different
steps in which the resistant elements withstand the seismic action.
The analyses performed allowed the identification of the main parameters affecting the structural behaviour of mixed building and gave indications on feasible modelling and verification criteria.
Fig. 20- Comparison between compressive stress – friction coefficient curves provided by PCI Handbook, CNR 10018 and
UNI-EN 1337:3 and tests regression curve.
Fig. 21- Proposed compressive stress - concrete–neoprene friction coefficient relationship.
4.5 TAMP: Influence of Infills on Structural Response
The objectives that have been pursued by this research are consistent
with the expected ones. Nevertheless, it must be observed that the
experimental calibration of the parameters defining the cyclic behaviour of the proposed equivalent diagonal strut model is affected by the
following main limitations: - it has been made considering only square
meshes of infilled RC frames; - the possible presence of an infill with
opening has not been considered.
These limitations, due to not sufficient time and resources, did not allow
the model to be validated by its use to reproduce the experimentally
detected response of the 3D infilled building subjected to the shaking
RESEARCH - Seismic behavior
table tests.
4.6 SCALE: Behavior and Strengthening of Stairs
The Task has pursued the proposed objectives. A detailed investigation
of the main stairs typologies and of the most used design procedures
have been performed; in particular, a report including the main results
was developed (Cosenza et al., 2007b). Numerical investigations have
been performed in order to understand the seismic behaviour, and the
possible failure mechanisms of buildings having the most common stair
typologies. The results have confirmed the need to utilize biaxial bending modelling and to account for the interaction of the different internal
forces as: bending moment-axial force interaction that characterizes the
inclined elements, and the bending moment-shear interaction that governs the behaviour of squat columns. Shear failure becomes predominant in the squat columns and in the reinforced concrete slabs and precedes the conventional ductile failure. An experimental set-up has been
designed on the basis of some simulations performed by using different
modelling: dimensions of a single span frame with inclined beam, loads
and resisting-wall have been defined.
4.7 NODI: Behavior and Strengthening of Beam-Column Joints
The research activities scheduled within the Task have been to a great
extent carried out without significant delays or anticipations, and the
main objectives have been achieved.
4.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
Summary of results reported in the previous sections leads to recognize
that the development of the work basically complies with the initial
schedule. In particular, as numerical analyses and software deliverables are considered, a good agreement with the program can be
In fact, different approaches and numerical procedures have been set
and made available to technical community. They cover at different levels the need of tools for checks required by modern codes in terms of
strength and local deformation. This applies both to design and assessment of existing un-strengthened concrete structures and to seismic
upgrading using FRP materials. Interaction between groups involved in
the study of irregular structures and of use of FRP for seismic upgrading of structures is another positive aspect of the work.
When experimental program are concerned, it is worth noting that a relevant contribution to the knowledge of bond mechanisms for smooth
bars has been given and an approach to the comparative analysis of performances in terms of rotation capacity of full scale r.c. members with
smooth and deformed bars has been accomplished. The work is not
RESEARCH - Seismic behavior
actually exhaustive, since trial biaxial tests have been designed depending on the findings of the theoretical work, but not completed. This
results in the need to extend and validate the results obtained in the
context of the present task and give a direct contribution to the development of specific design recommendations for members under combined axial load and biaxial bending.
4.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
The research activities scheduled within the Task have been to a great
extent carried out without significant delays or anticipations, and the
main objectives have been achieved.
5.1 MND: Non-Destructive Methods for the Knowledge of Existing
A large amount of RC buildings, both private and public, now placed
in seismic zones, were originally designed taking into account only
gravity loads and without explicitly provide ductile detailing. An
extraordinary rehabilitation program needs to be implemented on
such buildings, where an accurate evaluation of the available seismic
capacity is important to set up cost-effective interventions.
Investigations have a crucial role to adequately know the structure to
be evaluated. For this reason, there is an increasing need to set up
and put at disposal of technicians and other involved stakeholders
sufficiently reliable as well as not very expensive methods to estimate
in-situ material properties. Number of tests required to suitably apply
these methods have to be as low as possible, thus making the total
required budget sustainable to building owners and thus further
encouraging their use. To this purpose, the results obtained in this
Task confirm that a smart combination of NDTs and direct tests (such
as core extraction) gives effective solutions from both the economical
and technical point of view.
Future research work should be devoted to the following:
• Provide methods more and more capable of achieving effective
results in terms of prediction capability of concrete properties taking
into account both intrinsic randomness and epistemic uncertainty.
• NDTs currently available on concrete do not provoke damage on
structural members but on some other building components (e.g. partitions, infills, plaster, etc.) thus determining remarkable repair costs:
new methods are necessary to make them really not very expensive.
• As for reinforcement, taking into account the heavy damage caused
by the extraction of steel bars, non destructive methods to estimate its
mechanical properties need to be set up.
5.2 FC: Calibration of Confidence Factors
The work performed by the Task has focused on problems related to the
definition of a reliable assessment of structural seismic performance.
A probabilistic model for structural performance has been developed
and a method for calibration of Confidence Factors has been proposed.
Furthermore, a procedure for evaluation of material strength has been
developed based on the application of the Bayesian method, taking into
account, both, amount and reliability of the in-situ tests performed.
In addition to that, with reference to the prior probability distribution of
structural detailing parameters, a preliminary list of possible structural
defects was prepared, in which, for each defect, a set of possible values
and their relative weights are envisaged.
To proceed further along this path, the following developments are
• A complete procedure should be developed defining all phases of the
knowledge process of an existing building. The procedure shall give a
CF calibrated on the basis of the distribution of the assessment results
conditional on the acquired knowledge.
• Further studies are needed regarding the evaluation of material
strength values from in-situ non-destructive tests, whose reliability
depends on the reliability of the regression curves used to transform the
test parameter into a material strength value.
• A questionnaire, meant to be addressed to professional engineers as
a survey, has been prepared. The results of such survey shall be useful
in creating a thorough database of structural defects and their probability distributions.
5.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
Several directions for future work have emerged from the project, most
of them related to seismic code enhancements:
• Further studies for assessing the applicability of nonlinear pushover
procedures to plan irregular buildings;
• Further studies on different engineering demand parameters, such as
interstory drift, chord rotation, plastic hinge rotation, as means for
assessing the structural response;
• The need for the definition of the damping to be used in NTHA,
depending on the model level of refinement. Using 5% damping results
in an un-conservative assessment;
• A clear definition of spectrum compatible accelerograms (according
to EC8) to be used in NTHA;
• The need to establish a framework for Performance Based Earthquake
Engineering, based on a fully probabilistic approach;
• Assessment of the current modelling capabilities for shear failure prediction, with the possible development of simplified approaches that
can provide an engineering estimate without adding complexity to the
convergence procedure;
• Possible guidelines on how to consider other sources of nonlinearities/failures, such as bond-slip, beam-column failure, etc.
5.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
Considering the obtained results, the development of research aimed at
evaluating the seismic behavior of mixed-type buildings can be envisaged as follows:
• Establish through non-linear analyses the influence of RC-masonry
connections on the global behavior of the building. Particular attention
should be paid at intersections between beams and perimeter masonry
• Evaluation of q-factors. This objective should be pursued through the
implementation of (static and dynamic) non-linear analyses.
• Evaluation of the seismic response (numerical and not only) of other
structural configurations of mixed-type buildings not included in this
5.5 TAMP: Influence of Infills on Structural Response
This research gave useful results concerning the effects of masonry
infills on the lateral response of RC frames, supported by experimental
Possible developments could be aimed to pursue the following further
main objectives:
• to revise the available empirical expressions linking the masonry
properties to those of its components, by carrying out further tests on
masonry samples made of other kinds of resisting elements;
• to generalize the proposed model of equivalent diagonal strut, by evaluating experimentally the role of factors that have not been considered
here: infilled mesh geometry, different compression levels on the frame
columns after the masonry infills have been constructed, presence and
size of an opening in the infilled mesh of frame.
5.6 SCALE: Behavior and Strengthening of Stairs
Considering the obtained results the future research developing would
be oriented to the evaluation of the influence of the stair on the seismic
response analysing the following aspects:
• Experimental assessment of the strength capacity and deformability
of squat columns. In this way, the behaviour of squat columns would be
better understood considering the large number of models found in literature that are not completely exhaustive for a reliable assessment of
the seismic behaviour of reinforced concrete structures.
• Experimental evaluation of the seismic response of 2-D and 3-D
RESEARCH - Seismic behavior
frames with inclined elements and squat columns. The comparison of
different test results would allow to evidence the influence of the introduced specific elements such as inclined beams and squat columns.
• Numerical and experimental evaluation of the influence of the stair in
buildings with different location of infill walls. The numerical analysis
would be performed by using static and dynamic tools. This studies
would allow the evaluation of the interaction between stairs and infills
that are commonly modelled in different manner.
5.7 NODI: Behavior and Strengthening of Beam-Column Joints
The behaviour of beam-column joints can strongly affect the seismic
global behaviour of RC building structures. Some mechanisms (e.g.
concrete cracking, slippage of longitudinal reinforcing bars, etc.) can be
responsible of additional deformability, on one hand, while, on the other
hand, can alter the capacity design assumptions on the framing structural members (beams and columns) and on the joint member itself. For
this reason, research activities need to be increasingly devoted to develop accurate capacity models of beam-column joints to reliably evaluate
the performances of RC building structures.
Research carried out in this Task already provided important results
relevant to the role of the key behavioural parameters of RC beam-column joints, thus giving useful suggestions on the reliability of current
code expressions and on possible improvements. However, many other
issues need to be further studied regarding both as-built and strengthened joints (already damaged or not damaged), skilfully combining purposely designed experimental investigations, review of experimental
campaigns reported in the literature, and accurate numerical simulations.
Among others, some future research developments can be recognized as
• design and execution of extensive experimental programs on joint
specimens having different characteristics (e.g. internal or external, bior tri-dimensional, shape, beam type, etc.) well targeted on the types
representative of the Italian and European built environment;
• experimental and numerical validation of different strengthening
techniques based on the comparison of the relative performance and
application limits, particularly dealing with tri-dimensional joints and
joints with embedded beams;
• codification of test protocols to make possible and promote experimental result exchange.
5.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
This Task research activity dealt with a specific, but relevant aspect of
seismic design and assessment of r.c. constructions. An effective and
RESEARCH - Seismic behavior
fruitful approach to the development of tools for the theoretical estimation of strength and curvature ductility of members has been carried
out. Flexural mechanisms are clearly identified both at local and global scale and an encouraging capacity of simulation is demonstrated by
both static and cyclic proposed models of members with smooth bars.
This means that numerical sensitivity analyses can give a positive contribution to an optimized design of an experimental program able to
confirm numerical forecasts, show possible points of weakness of the
theories and/or give an insight on specific aspects of the behavior under
severe cyclic loads. The large variety of existing constructions and the
diffusion of smooth bars in many very urbanized areas exposed to seismic risk point out the need to continue the investigation on such type of
members and even develop customized strengthening techniques using
advanced materials.
Despite such positive feedbacks of the research, it is worth noting that
further work is strongly recommended to assess:
• the behavior of short columns, and
• the flexure-shear interaction in presence of smooth reinforcement.
In fact, the observed deformation mechanisms can produce effects on
the strength and ductility of members subjected to complex forces, like
those generated on columns of irregular constructions.
5.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
The experimental campaign and the numerical investigations carried
out have highlighted that the most vulnerable buildings are those with
disarticulated diaphragm behaviour. However, as emphasised in
Palermo et al. (2008), an accurate study on the modelling of connections needs to be done in order to correctly predict the overall response
of precast concrete buildings.
6.1 MND: Non-Destructive Methods for the Knowledge of Existing
- Masi A., Vona M. (2008). “La stima della resistenza del calcestruzzo
in situ: impostazione delle indagini ed elaborazione dei risultati”,
Progettazione sismica, Anno I, No. 1, IUSS Press, ISSN 1973-7432.
- Marano G.C., Morrone E., Mezzana M. (2008), “Approccio ibrido
fuzzy per l’integrazione e l’interpretazione delle prove non distruttive”,
Atti del convegno Valutazione e riduzione della vulnerabilità sismica di
edifici esistenti in cemento armato, E. Cosenza, G. Manfredi, G. Monti
editors, Polimetrica International Scientific Publisher, Roma, Italy.
- Olivito R.S., Spadea G., Carrozzini A., Spadafora A.R. (2008). “Strutture esistenti in cemento armato: controlli e verifiche mediante tecniche
non distruttive”. Atti del convegno Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in cemento armato, E. Cosenza, G.
Manfredi, G. Monti editors, Polimetrica International Scientific Publisher, Roma, Italy.
6.2 FC: Calibration of Confidence Factors
- Franchin P., Pinto P. E., Pathmanathan R., (2008). “Assessing the
adequacy of a single confidence factor in accounting for epistemic
uncertainty”, Convegno RELUIS “Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in cemento armato”, Roma 29-30
maggio 2008.
- Jalayer F., Iervolino I., Manfredi G., (2007). “Influenza dei parametri
di modellazione e dell’incertezza associata nella valutazione sismica di
edifici esistenti in cemento armato”, Proceedings XII Convegno ANIDIS, Giugno 2007.
- Jalayer F., Iervolino I., Manfredi G., (2008). “Structural modelling
uncertainties and their influence on seismic assessment of existing RC
structures”, submitted to Structural Safety, 2008.
- Monti, G., Alessandri S., Goretti, A., (2007). Livelli di conoscenza e
fattori di confidenza. Proceedings XII Convegno ANIDIS, Giugno 2007.
- Monti G., Alessandri S., (2008). “Confidence factors for concrete and
steel strength”, Convegno RELUIS “Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in cemento armato”, Roma 29-30
maggio 2008.
6.3 IRREG: Assessment of the Nonlinear Behavior of Buildings, with
Emphasis on Irregular Ones
- EuroCode 8 (2003a). Design of Structures for Earthquake Resistance,
- Part 1: General Rules, Seismic Action and Rules for Buildings, Draft
no. 6.
- EuroCode 8 (2003b). Design of Structures for Earthquake Resistance,
- Part 3: Strengthening and repair of buildings, Draft no. 3 (Final
Project Team – Stage 34), 2003.
- Fajfar P. (2002). “Capacity Spectrum Method Based on Inelastic
Demand Spectra”, Proceedings of the 12th European Conference on
Earthquake Engineering, Paper 843, London (UK).
- FEMA 356 (2000). Prestandard and Commentary for the Seismic
Rehabilitation of Buildings, Prepared by A.S.C.E., Washington, D.C.
- ATC 40 (1996). Seismic Evaluation and Retrofit of Concrete Buildings.
6.4 MIX: Assessment and Strengthening of Mixed-type (Masonry/RC)
- D.M. Infrastrutture, 14 Gennaio 2008. G.U. 4-2-2008, N.29, Norme
tecniche per le costruzioni.
- D.M. Lavori Pubblici, 16 Gennaio 1996. G.U. 5-2-1996, N.29, Norme
tecniche per le costruzioni in zona sismica.
- Decanini L.D., Liberatore L., De Sortis A., Benedetti S. (2006).
Rapporto sui comportamenti, osservati in laboratorio e dopo terremoti
severi, sulle prescrizioni di diverse normative, nazionali e internazionali, a proposito di edifici a struttura mista muratura-c.a.. Rapporto di
ricerca RELUIS. Novembre 2006.
- Decanini L.D., Tocci C., Liberatore L., Benedetti S. (2008). Analisi
numeriche non lineari agli elementi finiti su modelli piani e spaziali di
edifici a struttura mista muratura-c.a.. Rapporto di ricerca RELUIS.
Gennaio 2008.
- NAA-80, 1980. Normas Antisismicas Argentinas.
- Nardone F., Verderame G.M., Prota A., Manfredi G. (2008). Analisi
comparativa su edifici misti C.A.-Muratura. Valutazione e riduzione
della vulnerabilità sismica di edifici esistenti in c.a. Roma 29-30
Maggio 2008. MIX-04.
- O.P.C.M. n.3431, 2005. Suppl. Ord. n.85 alla G.U. n.107 del
10/05/2005 e s.m.i.
6.5 TAMP: Influence of Infills on Structural Response
- Amato G., Cavaleri L., Fossetti M., Papia M. (2008). “Infilled frames:
influence of vertical loads on the equivalent diagonal strut model”,
Proc.14th WCEE, Beijing, China, CD-ROM, Paper 05-01-0479.
- Cavaleri L., Fossetti M., Papia M. (2005). “Infilled frames: developments in the evaluation of the cyclic behaviour under lateral loads”,
Structural Engineering and Mechanics, Teckno Press, (21) 469-494.
- Decanini L.D., Bertoldi S.H., Gavarini C. (1993). “Telai tamponati
soggetti ad azioni sismiche, un modello semplificato, confronto sperimentale e numerico”, Atti 6° Convegno ANIDIS, Perugia.
- Durrani A.J., Luo Y. H. (1994). “Seismic retrofit of flat-slab buildings
with masonry infills”, Proceedings from the NCEER Workshop on
Seismic Response of Masonry Infills-Report NCEER-94-0004, 1-3.
- Mainstone R.J. (1971). “On the stiffness and strengths of infilled
frames”, Proceedings Institution of Civil Engineers, London,
Supplement IV, Paper 7360S, 57-90.
- Stafford-Smith B. (1966). “Behavior of square infilled frames”,
Journal of the Structural Division, ASCE, 92 (1), 381-403.
- Stafford-Smith B., Carter C. (1969). “A method of analysis for infilled
frames”, Proceedings Institution of Civil Engineers, London, (44) 31-48.
6.6 SCALE: Behavior and Strengthening of Stairs
- Berry R. (1999), The construction of buildings. 5th edition Vol. 2.
Blackwell Science.
- Biskinis D.E., Roupakias G.K, Fardis M.N., (2004).“Degradation of
RESEARCH - Seismic behavior
shear Strength of Reinforced Concrete Members with Inelastic Cyclic
Displacements. ACI Structural Journal title no. 101-S76.
- Cosenza E., Mariniello C., Verderame G.M., Zambrano A. (2007), Il
ruolo delle scale sulla capacità sismica degli edifici in C.A. Atti Del XII
Convegno Anidis “L’ingegneria Sismica In Italia”, 10-14 Giugno 2007,
Pisa, Italia. (in Italian).
- Cosenza E., Verderame G.M., Zambrano A. (2007), Rapporto sull’analisi tipologica delle scale e criteri di progettazione statica e sismica. Rapporto di ricerca RELUIS. (in Italian).
- Cosenza E., Verderame G.M., Zambrano A.(2008), Capacità sismica
degli edifici esistenti in C.A.: L’influenza delle scale, Proc. Reluis2rm08
“Valutazione e riduzione della vulnerabilità sismica di edifici esistenti
In C.A.” Roma 29-30 Maggio, E. Cosenza, G. Manfredi, G. Monti Eds.,
Polimetrica International Scientific Publisher, Isbn 978-88-7699-1255. (in Italian).
- Marrullier E. (1910), Guida Pratica per la costruzione degli edifici con
speciale riguardo al cemento armato, Editori Torinesi. (in Italian).
- Migliacci A. (1977), Progetti di strutture, raccolta delle lezioni tenute
presso il Politecnico di Milano negli anni accademici 1966-67, 196768, Parte seconda, Edizioni Masson, Milano. (in Italian).
- Pagano M. (1963), Strutture, Liguori Editore. (in Italian).
- Perri E. (1966), Ingegneria Sismica, Unione tipografico- Editrice
Torinese. (in italian).
- Reynolds C. E., Steeedman J.C., (2002). Reinforced concrete designer’s handbook, Spon Press Taylor & Francis Group London.
- Rosci L. (1939), Manuale pratico di volgarizzazione del Calcolo del
Cemento Armato, G. Lavagnolo, II edizione, Torino. (in Italian).
- Santarella L. (1953), Il cemento armato, le applicazioni alle costruzioni civili ed industriali, Vol. II, 13, Edizione Hoepli. (in Italian).
- Santarella L. (1957), Il cemento armato, monografie di costruzioni italiane civili ed industriali, Vol. III, 13, Edizione Hoepli. (in Italian).
- Sezen H., Moehle J.P., (2004). “Shear Strength Model for Lightly
Reinforced Concrete Columns, Journal of Structural Engineering,
ASCE November 2004, pp.1692-1703.
- Tecnica y Pratica del Hormigon Armado. (1989) Ceac ed. Barcelona.(in Spanish).
- Zhu L., Elwood K.J., Haukaas T., (2007). “Classification and seismic
safety evaluation of existing reinforced concrete columns”, Journal of
Structural Engineering, ASCE. 1316-1330.
6.7 NODI: Behavior and Strengthening of Beam-Column Joints
- Braga F., De Carlo G., Corrado G.F., Gigliotti R., Laterza M., Nigro D.,
(2001). “Meccanismi di risposta di nodi trave-pilastro in c.a. di strutture non antisismiche”, Atti del X Congresso Nazionale “L’ingegneria
Sismica in Italia”, Potenza-Matera, Italia.
RESEARCH - Seismic behavior
CEN Comité Européen de Normalisation (2003). Eurocode 8 – Design
of structures for earthquake resistance - Part 3: Assessment and retrofitting of buildings (draft n. 6)”, prEN 1998-3, Brussels.
- Masi A., Santarsiero G. (2008). “Sperimentazione su nodi trave-colonna in c.a. progettati con diversi livelli di protezione sismica: primi risultati”, Atti del convegno Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in cemento armato, E. Cosenza, G. Manfredi, G.
Monti editors, Polimetrica International Scientific Publisher, Roma,
- Masi A., Santarsiero G., Moroni C., Nigro D., Dolce M., Russo G.,
Pauletta M., Realfonzo R., Faella C., Lignola G.P., Manfredi G., Prota
A., Verderame G.M., (2008). “Behaviour and strengthening of RC
beam-column joints: experimental program and first results of the
research activity in the framework of DPC-Reluis Project (Research
Line 2)”, Proc. of the 14th World Conference on Earthquake Engineering,
Beijing, China.
- NTC 2008, Decreto Ministeriale 14 gennaio 2008, Norme tecniche per
le costruzioni, Suppl. or. n.30 alla G.U. n.29 del 4/2/2008.
6.8 BIAX: Behavior and Strengthening of Columns under Combined
Axial Load and Biaxial Bending and Shear
- Alessandri S., Monti G. (2008). “Design equations for the assessment
and FRP-strengthening of reinforced rectangular concrete columns
under combined biaxial bending and axial loads”. Journal Mechanics of
Composite Materials, 44 (3), pp. 197-324.
- Di Ludovico M., Verderame G.M., Iovinella I., Cosenza E. (2008a).
“Domini di curvatura in pressoflessione deviata di sezioni in c.a. –
PARTE I: Analisi a fibre”. Proc. Reluis2Rm08 “Valutazione e riduzione
della vulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International
Scientific Publisher, ISBN 978- 88-7699-125-5, pp. 629-638.
- Di Ludovico M., Verderame G.M., Iovinella I., Cosenza E. (2008b).
“Domini di curvatura in pressoflessione deviata di sezioni in c.a. –
PARTE II: Valutazione semplificata”. Proc. Reluis2Rm08 “Valutazione
e riduzione della vulnerabilità sismica di edifici esistenti in c.a.” Roma
29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., Polimetrica
International Scientific Publisher, ISBN 978- 88-7699-125-5, p. 639648.
- fib (2003) “Seismic assessment and retrofit of reinforced concrete
buildings”. Bulletin 24, ISBN 978-2-88394-064-2.
- Bosco M., Ghersi A., Leanza S. (2008). “Force displacement relationship for r/c members in seismic design”. Proceedings of the 14th World
Conference on Earthquake Engineering. October 12-17, Bejing, China.
- Canducci G., Camata G., Spacone E. (2008). “Problematiche connesse alla descrizione del moto sismico”. Proc. Reluis2Rm08 “Valuta31
zione e riduzione della vulnerabilità sismica di edifici esistenti in c.a.”
Roma 29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International Scientific Publisher, ISBN 978-88-7699-125-5,
pp. 327-334.
- Panagiotakos T., Fardis M.N. (2001). “Deformation of R.C. members
at yielding and ultimate”. ACI Structural Journal, 98 (2), pp. 135-148.
Park R., Paulay T. (1975). “Reinforced concrete structures”. J. Wiley
and Son, Inc., N.York, 769 pp.
- Verderame G.M., De Carlo G., Manfredi G., Fabbrocino G.(2008a).
“L’aderenza ciclica in campo elastico delle barre lisce. Parte I: la sperimentazione”, Proc. Reluis2Rm08 “Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E.
Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International
Scientific Publisher, ISBN 978-88-7699-125-5
- Verderame G.M., Ricci P., Manfredi G., Fabbrocino G. (2008b).
“L’aderenza ciclica in campo elastico delle barre lisce. Parte II: la modellazione”. Proc. Reluis2Rm08 “Valutazione e riduzione della vulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E. Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International Scientific
Publisher, ISBN 978-88-7699-125-5
- Verderame G.M., Ricci P., Manfredi G., Cosenza E. (2008c). “La
capacità deformativa di elementi in c.a. con barre lisce: modellazione
monotona e ciclica”. Proc. Reluis2Rm08 “Valutazione e riduzione della
vulnerabilità sismica di edifici esistenti in c.a.” Roma 29-30 maggio, E.
Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International
Scientific Publisher, ISBN 978-88-7699-125-5, p. 617-628.
6.9 PREFAB: Behavior and Strengthening of Prefabricated Industrial
- Felicetti R., Lamperti M., Toniolo G., Zenti C., (2008). “Analisi sperimentale del comportamento sismico di connessioni tegolo-trave di
strutture prefabbricate”. XVII Congresso C.T.E., Roma, 5-8 Novembre,
Italy, pp. 867-874.
- Magliulo G., Capozzi V., Fabbrocino G., Manfredi G., (2008). “Determinazione sperimentale del coefficiente di attrito neoprene-calcestruzzo per la valutazione della vulnerabilità sismica delle strutture prefabbricate esistenti”. Proc. Reluis2Rm08 “Valutazione e riduzione della
vulnerabilità sismica di edifici esistenti in c.a.”, Roma, 29-30 maggio, E.
Cosenza, G. Manfredi, G. Monti Eds., Polimetrica International Scientific Publisher, ISBN 978-88-7699-129-5, p. 717-724.
- Palermo A., Camnasio E. and Poretti M., (2008). “Role of Dissipative
Connections on the Seismic Response of One-Storey Industrial Buildings”, Proc. of the 14th World Conference on Earthquake Engineering,
Beijing, China.
he perception of the risk associated to the seismic vulnerability of
the transportation infrastructure, and in particular to that of bridge
structures, on the part of both the relevant authorities and the profession is a quite recent acquisition in Italy. This is possibly due to the fact
that in the last two major events that have struck the Country in the second half of the 20th century (Friuli 1976 and Irpinia 1980) the transportation infrastructure has not suffered significant distress. In particular, in Friuli the construction of highways was just at the beginning. In
the Apennine crossing of the A16 highway the bridges did undergo
some damage, mainly due to the inadequacy of the bearing devices, but
this was promptly remedied by the owner through the systematic adoption of the then innovative technique of seismic isolation.
On the other hand, it can be observed that this delay in the appreciation of the risk is not exclusive to Italy. For example, it is enough to
mention that it took twelve years after the spectacular failures of quite
modern bridges (Figure 1, left) during the San Fernando (1971) earthquake, for the Federal Highway Administration (FHWA) to publish a
first document titled “Retrofitting guidelines for Highway Bridges”
(FHWA-ATC, 1983). Still, in 1989, despite of the large retrofit program
set up (later proved to be fully inadequate), the Loma Prieta earthquake
exposed substantial deficiencies in bridges in California (Figure 1,
Fig. 1- Damage to bridges during the San Fernando, 1971 (left) and Loma Prieta 1989 (right) events.
The situation as briefly outlined above is sufficient to understand that
the state of the art on seismic assessment and retrofit of bridges still
needs to be advanced in several areas. The research undertaken within
this Line of the DPC-Reluis Project aimed at providing a contribution
in this direction. The areas considered to be of prioritary interest were
assessment methods, retrofit criteria and techniques, abutments and
foundations, with the final goal of producing a comprehensive document with guidelines and example applications. This result, which has
been achieved, represents the first European document on the topic and
RESEARCH - Seismic behavior
could be envisaged to form the basis for a future addition to the
Eurocodes system.
text on the seismic assessment and retrofit of existing bridges.
Starting from the year 1992 on funding from the FHWA, a vast research
program has been undertaken in the US to clarify several aspects related to the seismic assessment and retrofit of bridges.
The first product of the above research appeared in 1995 in the form of
the “Seismic Retrofit Manual for Highway Bridges” (FHWA, 1995): its
further development has led to the “Seismic Retrofitting Manual for
Highway Structures: Part 1 Bridges” and the “Seismic Retrofitting
Manual for Highway Structures: Part 2 Retaining structures, Slopes,
Tunnels, Culverts and Roadways” (FHWA- MCEER, 2005).
In Europe the Eurocodes system includes a normative document for the
seismic design of new bridges, which is at least partially based on the
recent concepts of performance-based design: Eurocode 8 Part 2 (CEN,
2005a). This document, however, is not matched by a companion one
covering existing bridges, differently with the situation of buildings, for
which such a document is available in the form of Eurocode 8 Part 3
(CEN, 2005b).
In the year 2003 a firm change of direction towards the harmonisation
with Eurocode 8 has occurred in the Italian normative framework for
seismic design. In that occasion the priority was given to the drafting of
documents for the design of new structures, both buildings and bridges.
A document for existing structures was also introduced, but again limited to buildings. These documents served later as the basis for the production of the seismic chapter of the current Eurocodes-aligned national design code produced by the “Ministero delle Infrastrutture”
The need for a document dealing explicitly with the problem of assessing and retrofitting bridges in seismic areas dates back actually to 2003,
when the update of the seismic design code was accompanied by the
obligation of assessing, within the time limit of five years, all the strategic structures and infrastructures in the Country. Adhering to this obligation and with reference to bridges, with funding from the Civil
Protection Department, the National Agency for Roads and Highways
(ANAS) has launched a program for the assessment of all its bridge
structures. Further, the theme is of pressing interest due to the widespread activity currently ongoing on the Italian highway network to
increase its traffic capacity.
Within the above context the DPC-Reluis research project, with its Line
3 is intended to respond to the outlined needs, and in particular to that
of producing a document to be used as a proposal of a future normative
RESEARCH - Seismic behavior
The program was articulated into five main tasks:
1) Identification of bridge typologies:
Bridge typologies characterising the Italian road- and railway systems
were identified through contacts with the main national and regional
administrations, as well as with major contractors. In particular, existing contacts were exploited and new ones were established with ANAS,
Autostrade, Italferr, RFI, Ferrovie della Calabria in order to acquire
detailed documentation on a number of representative structures.
2) Assessment methods:
Existing methods developed for the assessment of buildings were
extended to the deal with bridge structures, and methods specifically
devised for bridges were further developed. The goal was to fine-tune
several methods of increasing level of accuracy and required effort to be
used according to the importance and size/regularity of the bridge.
These include displacement-based linear and non linear static methods, as well as simplified behaviour-factor-based methods. Special
attention was also devoted to modelling for non-linear analysis.
3) Retrofit criteria:
This task focussed on the specific aspects of the application of traditional and innovative retrofit techniques to structural elements typical
of bridges. The program of activity included the execution of an experimental campaign aimed at establishing the effectiveness of alternative
retrofit techniques. This task also included the development of seismic
isolation solutions to be applied in the retrofit of bridges that were not
initially designed to be isolated, as well as bridge-specific seismic isolation design criteria.
4) Assessment of abutments, earth-retaining structures and foundations:
Abutments and foundations are often weak elements in existing
bridges. The goal of this task was to advance the state of the art in the
seismic analysis and assessment of these components, an area still
characterised by the widespread use of mostly empirical or conventional approaches.
5) Model applications to bridges of different typologies:
It was planned that under this task at least one bridge for each of the
main typologies identified under task 1 was to be subjected to a detailed
assessment and retrofit design, and then documented in an application
manual to complement the pre-normative document.
The guidelines and the companion manual represent the main
outcome of the project.
The five broad tasks outlined above were split into a number of sub33
Table 1. Subdivision of the research activity into tasks and sub-tasks.
tasks according to the table below. The table also shows the progress of
activity over the whole project duration of three years. The progress
achieved during the second year is briefly summarised in the following
Table 1.
(50+9×100+50 m and tall piers), Bardonecchia bridge (7×42 m) and the
Millaures bridge (6×80 m, composite steel-concrete). The second highway, built between 1965 and 1975 shows a greater typological variability, which can be reduced, however, to a few homogeneous sets.
Representative bridges are: the Borgotaro viaduct (slab bridge with several interconnections), the Narboreto bridge (4×30 m), the Rio Verde
viaduct (2?65+6×95+76 m and very tall piers, h=150 m) and the
Roccaprebalza South viaduct (13×45 m and tall piers), Rio Barcalesa
As it regards the infrastructure in the Abruzzi region typologies and
conditions were monitored along A14 and SS16. Data sheets, consisting of 7 sections that provide location, type, category, geometrical and
environmental characteristics, condition, and photographic description,
were compiled to catalogue all 52 bridges of the latter road. Bridges
were classified according to structural type, material and geometry. The
bridge conditions, the piers, and the cracking and vegetative maps were
The main results of the research activity are summarised in the following
according to the research structure described in the previous section.
4.1 Task 1: Identification of bridge typologies
The main bridge typologies on several seismic-prone portions of the
Italian railway and road/highway networks have been identified during
the first year of activity. In summary, the data collected from various
sources (mainly national/regional administrations) pertain to the TorinoBardonecchia-Frejus (TBF) and the Parma-La Spezia (PLS) highways
(Politecnico di Torino), the Firenze-Bologna (A1FiBo) portion of the A1
Milano-Napoli and the Apennine portion of the A16 highways (Università
di Roma “La Sapienza” and Università di Roma Tre), the Adriatic portions of the A14 Bologna-Canosa highway and of the SS16 Adriatica
state-road (Università di Chieti-Pescara), the Roma-Viterbo (RMVT) and
Roma-Sulmona (RMSu) railways (Università di Roma Tre), the regional
railway and roadway networks of Calabria (Università di Cosenza).
Structural typologies characterizing the TBF and the PLS highways are
quite different. The first highway, built in between 1983 and 1992,
includes rather uniform typologies: a) about 300.000 m2 of precast segmental box girder bridges with pier heights up to 90m and span lengths
between 40 and 100m, b) about 200.000 m2 of girder bridges in concrete and in composite steel-concrete with pier heights between 5 and
30m and span lengths between 20 and 80m. Representative bridges
are: the Borgone viaduct (20+26×40+20 m), the Ramat viaduct
Bridge structures on the A1FiBo and A16 were scrutinised, and a
selected number of bridges either representative of the most frequent
typologies or significant for their outstanding design was identified. A
further screening of the set including these structures and those identified on the TBF and PLS has led to the definition of the final case-studies for the detailed applications and the development and calibration of
assessment methods.
In the roadway and railway system of Calabria the most common typologies are single-stem or frame piers with full, single-, or multi-cellular
hollow-core cross-section; simply supported decks, made up of reinforced or prestressed concrete girders and a cast-in-place RC deck
slab. Two study cases were selected for the study of non-conventional
protection/retrofit techniques. The first one is the Follone viaduct on the
A3 Salerno-Reggio Calabria highway, where the spans have been connected with by means of longitudinal devices, while the second case is
the Val di Leto viaduct, on a provincial road, which was recently retrofitted using oleodynamic devices.
Finally, the data collected during the survey of the two railway lines
RMVT and RMSu, have allowed selection of four masonry arch bridges,
two per line, to be used as case-studies for the calibration of analysis
methods for masonry bridges.
4.2 Task 2: Assessment methods
In summary, Task 1 has shown that the relatively many important
RESEARCH - Seismic behavior
bridges crossing wide valleys in the mountain tracts of the Central and
Southern Apennine (A1FiBo, PLS, A3) represent a negligible percentage of the total bridge stock, made up essentially of bridges with simply-supported decks (prestressed or reinforced concrete girders plus
slab) with single stem of frame piers. For this reason a special effort has
been devoted to devising a simplified non-linear method suitable for the
analysis of bridges with simply-supported decks (Università di Roma
La Sapienza). As it regards statically indeterminate bridges (continuous
decks) a distinction was made between those with special configuration,
for which inelastic dynamic analysis is in most cases the method of
choice, and simpler bridges, for which two methods have been thoroughly explored: the Modal Pushover Analysis (MPA) method
(Università di Roma La Sapienza) and the Secant Mode Superposition
(SMS) method (Università di Pavia). Finally research has also focussed
on two more issues, namely the always debated problem of directional
combination rules (Università di Chieti-Pescara) and the non linear
modelling of seismic protection devices (Università di Cosenza).
m = 0.3mpila + mpulv + mimp
(m + 0.3mpila)Hp + mimpHimp
H pulv
y = 1 yH2 / 3
u = y + (u – y) lp(H–lp / 2)
T = 2 mi / k = 2 miy / Vy
max= SDe(T)
max= SDe(T) 1+(q–1) Tc
T Tc o q 1
T < Tc
The guidelines and application manual give a detailed description of
the method, of the safety verifications of bearings, piers and foundations, and a complete worked-out example.
For these bridges it is possible to set up an ad hoc assessment procedure which represents a convenient trade-off between simplicity and
accuracy. The reference model is that of a vertical cantilever with a continuous distribution of mass, on top of which rest the pier cap and the
deck masses. As long as the pier height is not such as to make higher
mode contributions significant, in the transversal direction each pier
represents a single-degree of freedom oscillator (see Figure 2a). In the
longitudinal direction the entire bridge can also be represented as a
SDOF system if seismic restrainers are provided that minimise the relative movements of adjacent decks on top of the pier caps (see Figure
2d). In this case the system has mass equal to the sum of the tributary
masses of the piers and resisting force sum of the resisting forces of the
piers (assuming that maximum displacements are permitted by the
abutments joints).
The method consists of a simplified non linear static analysis in which
the force-displacement laws are constructed based on the results of
moment-curvature analysis of the pier bases (see Figure 2b). The following equations give the tributary mass, Eq.(1), the effective height in
the transversal direction, Eq.(2), the yield and ultimate displacement
(see Figure 2b and c), Eq.s (3) and (4), the period, Eq.(5), and the corresponding demand displacement, Eq.(6), as for single-mode conventional pushover analysis. The effective height equals the pier’s height in
the longitudinal direction. Displacement capacity follows from ultimate
displacement with appropriate safety factors.
RESEARCH - Seismic behavior
The MPA method by Chopra and Goel (2002) has been devised for the
analysis of tall buildings. Its applicability as an alternative to inelastic
dynamic analysis and adaptive pushover methods for the assessment of
Fig. 2- Simplified non linear method for the assessment of bridges with simply-supported decks.
Fig. 3- Longitudinal profile of the Rio Torto viaduct (A1FiBo).
• The variation of the lateral load distribution, from one based on the
modal (elastic) displacement shape to another based on the (plastic)
displacement shape at failure, does not affect appreciably the results.
• The best estimate of the displacements by the MPA (i.e. the one
derived taking the optimal reference DOFs) matches reasonably well
that from TH. It is worth noting that a comparable amount of approximation on the response of the structure is obtained both in the elastic
and in the plastic response regimes. This observation, together with the
previous one, indicates that the main approximation of the method, i.e.
being based on the initial elastic modal vector, may not represent a
major limitation.
• Differences between the nodal displacements estimated by the MPA
with respect to those by the TH are found to be in the order of 15%,
independently of the intensity level of the ground motion. Analogous
results are observed also for the curvatures at members end-sections,
resulting in almost coincident patterns of plastic hinge location and predictions of members failures.
Fig. 4- Two piers of the Rio Torto viaduct (A1FiBo).
bridge structures has been investigated through its application to the Rio
Torto viaduct (see Figures 3 and 4), one of the case studies selected for
the project. The structure, built at the end of 50’s, is characterized by
thirteen-span twin decks realized with two girders and top slab. The
twelve supports consist of a pair of framed piers, one under each deck.
Each pier is a multi-storey reinforced concrete frame with variable
height, realized with two circular columns of diameter D =120÷160 cm.
The results from inelastic dynamic analysis have been taken as benchmark for the purpose. The response has been compared for several
intensity levels (to assess the influence on accuracy of the level of nonlinearity in the response) and in terms of different response quantities,
both local and global (section curvatures and element displacements).
The comparison provided the following indications:
• The location of maximum modal displacement is the best choice as
the reference degree of freedom (DOF) for estimating the demand on the
structure. Each significant mode is therefore characterized by its own
reference DOF.
For the considered case, the application of the MPA method has shown
to lead to fully acceptable results. Such a favourable conclusion still
awaits substantiation from a larger number of applications. These
results have led to the introduction of the method among the allowed
methods in the draft guidelines.
The Secant Mode Superposition method consists essentially of an iterative multi-modal response spectrum analysis on a structural model
with secant stiffness properties and equivalent viscous damping. The
procedure can be summarised in the following steps:
• Step 0: A starting displacement profile and stiffness distribution are
• Step 1: The stiffness matrix of the equivalent linear structure is
• Step 2: Modal analysis is carried out;
• Step 3: Displacement in each vibration mode are obtained either from
an over-damped elastic or from an inelastic displacement spectrum;
• Step 4: Modal contributions are combined to yield displacement proRESEARCH - Seismic behavior
file and moment distribution (different combinations rules were examined);
• Step 5: Two response indices are computed, that evaluate convergence on displacement profiles and force distributions, while checking
that the structural capacity is not violated.
• Step 6: A final response index is obtained as an average of the first
two and checking the convergence of the proposed iterative procedure.
The method has been thoroughly tested on the six “typological” bridges,
with regular and irregular configurations, and different number of spans
and span length. Verification of the method is versus non-linear timehistory analysis in terms of maximum deck displacement, and maximum pier shear forces has been carried out.
4.3 Task 3: Retrofit measures
The experimental part of the research activity of Line 3 has been carried out at the University of Pavia and of Roma Tre. The two experimental campaigns have focussed with different goals on the testing of
piers. The tests performed in Pavia were aimed at ascertain the effectiveness of FRP retrofit measures in order to confine hollow-core piers
with insufficient lap-splices, while those performed in Roma Tre were
aimed, using large-scale specimens, at the characterisation of the
response of frame piers built in the ‘60s.
Finally, as it regards masonry bridges, a comprehensive survey of the
existing retrofit techniques for this type of structures has been carried
out by the University of Genova.
Four 1:2 scaled bridge piers were designed with an insufficient overlapping length of the longitudinal bars across the critical zone that
should lead to an early loss of the lateral strength due to bar slippage.
The built specimens (see Figure 5 left and middle) have the following
• Hollow-core rectangular cross-section (see Figure 5, right) with external dimensions 800x1500mm and wall thickness of 150mm;
Table 2. Considered retrofit materials
SRP 3x2
High Density
SRP 12x
High Density
High Modulus
Alkali Resistant
High Modulus
RESEARCH - Seismic behavior
Fig. 5- Pier section and specimens built at the University of Pavia.
• Pier height of 6 m (aspect ratio equals 4);
•Longitudinal reinforcement: 8010 (rL = 1.05%) with an overlapping
length equal to 20 diameters (200mm) at the base of the pier;
•Transversal reinforcement: stirrups 6/150mm (rV = 0.38%);
• Axial load equal to 1000kN ( = 4.3%) or 2000kN ( = 8.6%);
• Concrete Rck400;
• Steel FeB44K.
The retrofit intervention aimed to restore the tensile stress path from the
pier section to the foundation, avoiding at the same time any plasticization of the overlapping region. The new stress path created using longitudinal FRP strips applied to the overlapping region is expected to
cause the plastic hinge shift upwards where the longitudinal steel is
well anchored allowing for an efficient energy dissipation.
During the design phase different possible solutions have been considered concerning the retrofit materials (carbon, aramid or glass FRP), the
retrofit geometry (width and length of the region to be retrofitted), the
techniques to be used for the anchoring of the FRP strips to the foundation. This was possible employing a numerical FE model developed
to predict/reproduce the tests results.
Regarding the materials, the final choice was to use carbon FRP (CFRP): the analyses indicated that this material is the only one able to
sustain the acting tension forces. Too many FRP layers would have been
needed to carry the same force using glass or aramid fibres, affecting
the effectiveness of the retrofit intervention.
For what concern the geometry of the retrofit intervention, the final
solution was to apply longitudinally two C-FRP layers on the four sides
of the specimen. As far as the exploitation of the material strength is
Layer thickness [mm]
concerned this choice appears to be questionable since the fibres
applied to the pier sides parallel to the imposed motion will not have
the same stress as those on the other two sides, but the adopted solution
seemed to be the only possibility to assure the maximum stress diffusion across the pier section. It is worth mentioning that even though
anchoring 1500kN force to the foundation of the scaled specimen would
have been probably feasible using a steel collar fixed to the foundation
with some high-strength steel bars taking advantage of the deep foundation of the specimen, moving back to real structures the anchoring to
the foundation of the tensile force induced in the FRP by a seismic
excitation would have been much more difficult, if not unfeasible.
Spreading the tensile force on the four sides of the pier, the anchoring
is clearly easier. Between different possibilities initially considered to
anchor such force, final choice was to use an anchoring system realised
with FRP too. The idea was to employ aramid connectors, normally
used to transfer shear stresses. If this solution will be found to be effective, as it seems from its design, multiple advantages will arise both on
the economic and technologic sides.
Due to external constraints only two piers have been tested within the
duration of the project, those without the FRP retrofit in the lap-splice
region. The tests confirmed that, as expected, lap-splice with an overlapping length equal to 20 times the diameter of the spliced bars is
insufficient to assure the anchoring of the bars (see Figure 6a). The tests
also underlined that the effectiveness of the lap-splice decreases while
the axial load increases: that is because of the higher stresses and damages (such as partial concrete spalling) in the overlapping region.
Figure 7 shows the base shear-top displacement diagrams derived from
the performed tests: here the lateral load carrying capacity drops quite
quickly because of the bars sliding. The red curves are the result of the
numerical model.
Given the large number of seismically under-designed bridges, that
need to be assessed and potentially retrofitted due to insufficient lapsplicing, the development of an efficient analytical model to simulate
the response of FRP-retrofitted elements was deemed critical. A quite
simple though effective finite element model was developed using
Seismostruct (Seismosoft, 2006). Figure 8 shows the adopted numeric
model. The longitudinal FRP layers have been represented like an element itself. Rigid links have been used to place the FRP at the right
distance from the longitudinal axis of the retrofitted member in order for
them to be able to give the right contribution to the flexural strength of
the pier. Furthermore, each FRP element has pinned connection at both
ends in order to be subjected to pure axial load. On the other hand, the
FRP wrapping can be modelled in approximation without adding elements to the model, since its main effect is the increased concrete confinement that can be represented by the confinement factor already present in adopted concrete stress-strain representation (Mander, 1988;
Martinez-Ruenda and Elnashai, 1997). To tests the effectiveness of the
Fig. 8- FRP retrofitted pier model.
Fig. 6- Test Set-up (a) and open crack at the pier base during the test with 2000kN axial load (b).
adopted finite element model, the behaviour of 1:4 scaled square hollow section piers from previous experimental campaigns (Calvi et al.,
2005 and Pavese et al., 2004) has been reproduced through push-over
Fig. 7- Base shear-top displacement diagrams of the two “as-built” pier (a) N = 1000kN (b) N = 2000kN.
Large-scale tests on framed piers have been undertaken at the
Università of Roma Tre. This typology, characteristic of many old
viaducts of the Italian highway system, has been chosen for its high
RESEARCH - Seismic behavior
Fig. 9- The viaduct “Rio Torto”.
beam, (see Figure 11a), a suitable grid of displacement transducers has
been placed on this beam, in order to measure the cracks amplitude, in
the other two specimens. In the second two mock-up’s a different failure mechanism occurred on the same transverse beam. In particular,
during the second test both beam-column joints collapsed (see Figure
11b-c), while in the third one, only the left end of the beam failed in
shear (Figure 11d), with a simultaneous failure of the right beam-column joint (Figure 11e). This outcome was a nice experimental verification of the effect of material fluctuation on determining which amongst
similarly resistant failure mechanisms actually occurs in reality.
Fig. 10- The viaduct “Rio Torto”.
seismic vulnerability. Among the representative bridges, a framed pier
from the “Rio Torto” viaduct has been chosen (see Figures 3, 4 and 9).
For the experimental program three mock-up’s of pier 12 without retrofit have been realized and tested, with the goal of characterizing its
cyclic response and relative collapse mechanism (see Figure 10).
Subsequently, one or more reinforcing systems were meant to be
applied to the tested piers, for repeated testing to check the efficiency
and the reliability of the proposed reinforcing solutions.
A ductile flexural failure was predicted with formation of plastic hinges
for this pier while, on the contrary, all three specimens failed in shear,
either in the transverse beam or the joints, suggesting that the formula
employed for the evaluation of the shear strength tends to overestimate
the ultimate shear.
Since the first test has shown a premature shear failure of the transverse
Fig. 12- Experimental force-displacement cycles.
Fig. 11- Failure mechanisms of the transverse beam in the three tests.
RESEARCH - Seismic behavior
4.4.1 A
Fig. 13- left: Comparison between theoretical and numerical force-displacement curves; right: base column rotations.
The differences between the failure mechanisms of the three piers, however, have a little influence on the global behaviour, as shown in Figure
12, which compares the global force-displacement cycles of the three
The experimental results have been compared with the results of a
numerical model, which was set up using the non-linear code
“OpenSEES”. Shear failure has been introduced using a shear forcedeformation relationship with a tri-linear backbone and an appropriate
degradation law, included in a fiber non-linear element, using the section aggregator command. The yield-penetration at the base of the column effect is particularly relevant due to the presence of plain steel
bars. This phenomenon, if neglected, can induce an overestimation of
the structural stiffness. This effect has been taken into account using a
zero-length element placed at the column base with a properly modify
stress-strain law of the steel bars. Finally, the buckling phenomenon
has been taken into account using a corrected constitutive law of steel.
The FE model used was able to reproduce accurately both the global as
well as the local behaviour, as shown in Figure 13.
A simplified model for the dynamic analysis of diaphragm walls retaining dry cohesion-less soils with horizontal back-slope subjected to seismic excitation has been developed (Franchin et al, 2007a). The model
is based on the well-known one-dimensional Winkler approximation
and on the non-linear shear-beam model for the ground layers on both
sides of the wall (see Figure 14). The model can include anchor-ties and
can account for non-linearity in all of its elements (retained soil,
anchors and wall). According to preliminary numerical applications,
which include validation of the proposed model results versus those of
a refined plane-strain nonlinear finite-element analysis carried out with
a commercial code, the model appears to yield quite accurate predictions of static and dynamic bending moment distributions and permanent wall displacements.
Next the developed model has been applied for the analysis of the
response of a diaphragm abutment prior and after upgrading intervention with change of the support conditions and insertion of tie-backs
(Franchin et al 2007b). The analysed structure is represented in Figure
The application of the model has shown its versatility in assessing the
system response in its existing state and in progressive states of upgrad-
4.4 Task 4: Abutments and foundations
The activity under this task has been carried out at University of Rome
La Sapienza, and has dealt with two distinct problems: a) the development of an efficient non linear method for the analysis of diaphragmtype abutments, free standing and retrofitted with tie-backs; b) the
review of the literature on soil-foundation-structure interaction with the
goal of providing detailed indications for practitioners to be included
into the assessment guidelines.
Fig. 15- Diaphragm abutment retrofitted with anchor ties.
Fig. 14- The developed model for diaphragm type.
Fig. 16- Results of the abutment analysis: left, moment diagrams; right, top displacement time-histories.
RESEARCH - Seismic behavior
A comprehensive review of the literature on the treatment of the
response of deep foundations has been carried out. This has led to identifying the available methods and their pros/cons. After the survey a
selection has been made of those procedures considered suitable for
practical application and some numerical applications have been carried out to assess the relevance of the phenomenon (input motion modification by kinematic interaction and foundation flexibility), in terms of
the response of the superstructure.
One example is the bridge structure shown in Figure 17. It is a simplysupported prestressed concrete deck of span length 30.0m typical of the
Italian highway construction practice of the ’50s-‘70s with piers consisting of a single-stem with hollow-core circular section. The dimensions are in the figure. The foundation consists of a mat on 5 piles of
soil-foundation system, consisting of the stiffness and damping functions (of the frequency) to be assigned at the pier base. This impedance
includes the evaluation of the frequency dependent “dynamic” group
effect, i.e. the modification of the impedance obtained as a simple summation of the individual pile impedances to account for the interaction
of the wave-fields produced by each pile.
– Evaluation of the response in the frequency domain. This has been
done both with a purpose-made code and with a commercial finite element software that implements frequency-domain analysis (Sap2000).
The resulting power-spectrum of the displacement components can be
integrated to yield the root-mean-square (RMS) or standard deviation of
response, from which maxima to be used in verification are readily
ing, in terms of both forces (Figure 16 left) and dynamic displacements
(Figure 16 right). To the extent that it has been validated at present, the
model represents a very efficient tool for realistic design and assessment purposes.
Fig. 17- Simply-supported deck on single-stem hollow-core pier founded on piles.
Fig. 18- Complex impedance at the pier base: stiffness (top), camping (bottom), translation (left) and rotation (right).
1.5m diameter. Pile length is 20m. The bridge has 6 spans and a pier
of height 20m has been considered. Soil can be classified based on the
available information as type D. The structure has been modelled as
shown in Figure 17e, i.e. as a three-degree of freedom system (including horizontal and rocking component of the base).
The analysis has been carried out in the frequency domain using the
substructuring approach. The steps of the analysis include:
– Evaluation of the modification of the surface free-field motion (supplied as an acceleration response spectrum) due to the kinematic interaction between soil and pile group. This step provides the power spectrum of input displacement at the pier base.
– Evaluation of the complex frequency-dependent impedance of the
obtained by multiplication for the peak factors.
RESEARCH - Seismic behavior
Figure 18 shows the real (stiffness, top) and imaginary (damping, bottom) parts of the complex impedance at the base of the pier, for the
translation (left) and rocking (right) displacement components. These
are reported for two different values of the shear wave velocity Vs, both
compatible with the soil type D. The figure also reports the stiffness/damping obtained by simple summation of the single pile contributions. Comparing the latter with those of the group allows to appreciate the frequency-dependent effect of the pile-to-pile interaction. This
effect reduces, by more than 50% in this case, the total stiffness.
Finally, Figure 19 shows the power spectral densities of the response in
Fig. 19- Results of SSI analysis on a bridge pier: power spectral densities of the response in terms of total displacement (left)
and structural deflection (right).
terms, on the left, of total displacement (relative to input motion, i.e.
sum of the foundation translation, the structure deflection and the translation due to rigid foundation rotation), and on the right of the structural deflection only. Results are reported for the two Vs values and, for
reference, for the fixed-base response. As it can be seen, as expected,
the fundamental period of the system elongates considerably due to the
introduction of the foundation flexibility: it starts at T=0.83 s in the
fixed base case, and reaches about 1.5s and 1.75s for Vs=200m/s and
100m/s, respectively. This increases the total displacements. The drifts,
however, are considerably reduced as shown in Figure 19b.
The above application, as well as the others carried out, allowed to
introduce in the guidelines quantitative indications on the need for
inclusion of SSI into the modelling.
4.5 Task 5: Numerical application to case-studies
All research units have contributed in producing a vast amount of casestudies that have been of considerable usefulness in checking consistency and practicality of the indications that now form the guidelines for
assessment. In this section only a limited overview of the applications
is given to illustrate the work done. A more detailed description can be
found in the final report for the Line 3. Table 3 reports all the analysed
4.6 Guidelines and Application manual
The activity only briefly summarised in the previous sections has represented a necessary support for undertaking the task of writing what
was the final product expected from Research Line 3: a proposal for a
guidance document on the seismic assessment of existing bridges, and
a companion set of example applications. The task, carried out by
University of Rome La Sapienza, has gone through several rounds of
scrutiny by all the units. In its final version it represents the first
European document on the topic and could be envisaged to form the
basis for a future addition to the Eurocodes system. Indeed, the document is fully in line with Eurocodes and reflects to some extent the
experience on the seismic assessment of existing structures gained with
the use of Eurocode 8 Part 3 on buildings. It is also in line with the relevant chapters of the DM2008, related to seismic design of bridges, and
incorporates its most recent developments on the definition of seismic
The document produced consists of four chapters and two appendices:
• Chapter 1: gives an introduction to the problem of seismic assessment
of existing bridges;
• Chapter 2: contains the guidelines;
• Chapter 3: is an overview of the most common retrofit measures and
criteria employed, without entering into the specifics of their design,
making reference for this purpose to specialised texts on the topic;
Table 3. List of case-studies analysed according to the assessment guidelines
4 simply supp. spans, circular hollow-core piers
Narbareto (PLS)
7 simply supp. spans, polygonal bi-cell. piers
Rio Barcalesa (PLS)
Hollow-core slab deck, highly irregular plan, Figure 21
Borgotaro (PLS)
9 spans, steel deck with hollow-core RC piers
Rio Verde (PLS)
5 spans, box-section, steel pier
Ramat (TBF)
15 spans, frame piers
Vasto Marina (SS16)
10 spans, box-section, single-stem hollow-core piers
Della Valle (A25)
13 spans, inelastic time-history analysis
Roma Tre
Rio Torto (A1FiBo)
13 spans
La Sapienza
Rio Torto (A1FiBo)
spans, simply-supported, and continuous after section
La Sapienza
Standard viaduct (E45)
4 simply supp. spans, retrofitted with “link system”
Follone (A3)
5 simply supp. spans, retrofitted with oledynamic devices
Val di Leto
Elastic RS analysis + q-factor
Elastic RS analysis + q-factor
Elastic RS analysis + q-factor
Elastic RS analysis + q-factor
Elastic RS analysis
Elastic RS analysis + q-factor
Elastic RS analysis + q-factor
Inelastic time-history analysis
Modal pushover analysis
linear dynamic
Inelastic time-history analysis
Inelastic time-history analysis
RESEARCH - Seismic behavior
Fig. 20- The Narbareto viaduct.
North carriageway
South carriageway
Fig. 21- The Borgotaro viaduct.
• Chapter 4: contains the numerical examples that illustrate the application of the methods presented in the guidelines. There are four applications covering:
– assessment, by means of the simplified non linear method, of a typical simply-supported bridge with single-stem cantilever piers in its present state;
– assessment, by means of linear and pushover analyses, of the previous bridge in two different configurations, with a new continuous, wider,
RESEARCH - Seismic behavior
composite steel-concrete deck, with and without seismic isolation;
– assessment of the Rio Torto viaduct by means of inelastic time-history analysis.
• Appendix A: presents the fundamentals of the response to multiplesupport excitations and reviews a number of methods that can be
employed to analyse bridge structures for this effect;
• Appendix B: presents the fundamentals of the soil-foundation-structure interaction phenomenon and reviews a number of methods that can
Fig. 22- The Della Valle viaduct.
Fig. 23- The Follone viaduct.
RESEARCH - Seismic behavior
be employed to analyse bridge structures for this effect.
The main body of the manual is represented by chapters 2 and 4, as
well as by the appendices. In the following the most significant or problematic aspects are briefly reviewed and commented.
The seismic action is defined, in line with DM2008, by means of an
elastic acceleration or displacement response spectrum characterized
by an average return period specified as a function of the limit state of
The return period TR is obtained from the probability of exceedance
PVR over the reference life VR. The latter is given in DM2008 as the
product of two factors, the nominal life VN and the “use factor” CU. The
minima for PVR for each limit state are given in DM2008.
In the tentative applications of the guidelines it was raised the problem
of the value to be attributed to VN and CU, especially with reference to
the first one. The uncertainty may arise in the choice between 50 and
100 years for VN, when considering bridges over highways. The
DM2008 indicates 50 years for bridges of ordinary dimensions, typology and importance, and 100 years for bridges of large dimensions and
“strategic” importance. One would then be probably directed towards
100 years, in consideration of the importance of the bridge (it is on a
highway). The next choice is that of CU which leads unambiguously to
2.0, since highways are roads of type A according to the Italian classification of roads (i.e. considering, again, the functional importance of
the road on which the bridge is located). The above choices would imply
a reference life of 200 years and, for the life-safety limit state, a TR of
about 2000 years. It is observed that this conclusion would not to be in
line with the safety criteria contained in Eurocode 8 Part 2 (Bridges)
which indicates for highway bridges an importance factor I=1.3 to be
applied to the action with TR = 475 years. This multiplication leads in
most of Italy to an action with a return period of about a 1000 years.
This latter in turn is consistent with a reference life of about a 100
years, which is also the design life specified in the Eurocodes for other
actions (e.g. corrosion).
the possibility of defining the concept of “residual” reference life.
Though it is admitted that in our Country it seldom occurs that the decision to demolish a bridge can be taken several years in advance, it may
happen that, due to planned substantial modification of the traffic
capacity of the link, it will be economically more convenient at a future
date to replace the bridge. In this case, if seismic upgrade must be
undertaken, the concept of residual reference life may be invoked to
assign to VR a more realistic reduced value. This possibility is not currently included in the guidelines, though it is regarded as being in line
with the possibility allowed for existing structures to derogate from
standard safety levels dictated for new structures.
With respect to classification of methods in static and dynamic, linear
and non linear, now common to all modern seismic design codes and
giving rise to the usual four alternatives, the guidelines restrict somewhat the field of applicability of linear analysis. This is not unexpected.
For new well-designed structures the role of analysis is a relatively
minor one, due to the many constraints (arising mainly from global and
local capacity design) that guide the design. On the other hand, when
assessing an existing structure, the accuracy in the analysis may have
a major economic impact on the retrofit, possibly avoiding it altogether.
An official response to the mentioned problem, whose relevance needs
not to be underlined, cannot but come from the competent authorities,
which are in charge of choosing the safety levels.
The guidelines admit linear analysis of two types only: modal analysis
with unreduced elastic spectrum and verifications in terms of deformation/forces (subject to stringent conditions on the response regularity),
and modal analysis with a spectrum reduced by a limited value of the
behaviour factor of q=1.5.
The main methods put forward by the guidelines are non linear static
and dynamic analyses. As already anticipated in § 4.2.1, a simplified
non linear static method is proposed for the very frequent case of
bridges with simply supported decks. For continuous irregular bridges
the use of more recent pushover variants (adaptive and/or multi-mode)
is introduced as an alternative to full-fledged inelastic time-history
analysis. The allowance for more than single-mode invariant pushover
represents a small step forward with respect to Eurocode 8 Part 2,
which builds upon the results of recent wide-ranging studies on the performance of such methods in the analysis of bridges [see for ex.
(Casarotti 2005), (Kappos et al, 2005), (Isakovic and Fischinger, 2005),
(Lupoi et al, 2007), (fib, 2007), as well as the draft document “Inelastic
methods for seismic design and assessment of bridges” by Task Group
11 of the European Association of Earthquake Engineering].
Within the framework of the definition of the reference life one aspect
that deserves particular consideration in the case of existing bridges is
The guidelines introduce a format for bi-directional verification for both
RESEARCH - Seismic behavior
deformations and forces. In particular the format reads:
Dx 2 Dy 2
Cx Cy
)( )
where Dx and Dy denote the demand quantities along the two orthogonal axes x and y, with Cx and Cy denoting the corresponding capacities.
This format becomes, in terms of chord rotations and shear forces:
(x/u,x) + (y/u,y) 1
(Vx/Vu,x) + (Vy/Vu,y) 1
In the above equations the demand terms are understood as the combined effect of both orthogonal components of the seismic action. For
example, with reference to chord rotation, for the case of multi-modal
non linear static analysis one has:
x=xG± [(
xEx,i–G) + (xEy,i–G) ]
where the directional combination is of the SRSS type and the summation is over the modes.
dent motions at the supports representative of the local soil conditions,
which can be applied using currently available commercial finite element codes (Monti and Pinto, 1998);
For what concerns soil-foundation-structure interaction the guidelines
give a classification of the approaches and present with some detail the
substructuring method, in its application to pile (Novak 1974, Makris
and Gazetas, 1991 and 1992) and caisson foundations (Gerolymos and
Gazetas, 2006a,b). In this method the structure and the soil-foundation
system are separated and studied accordingly. The study of the soilfoundation system consists of the solution of so-called kinematic interaction and inertial interaction problems, leading to the modified input
motion for the structure and to (complex) impedance to be put at the
structure base, respectively. Then the structure is analysed, with a flexible support condition, under the previously determined modified
motion. All the formulas necessary to perform this procedure are present in the Appendix.
The matter covered in these two appendices, i.e. the response of bridge
structures to different motions at the piers’ bases and the effect of the
soil-foundation system deformability in modifying the input motion as
well as the response of the structure, has been always mentioned in
codes without, however, neither precise quantitative indications on the
instances in which these phenomena have to be accounted for, nor of
physically sensible yet practically applicable methods to do it. The reason for this resides clearly in the insufficient advancement on basic
research. In drafting the guidelines, however, it was considered appropriate to provide a presentation of selected state-of-the-art approaches
which are susceptible of practical application.
For what concerns the effect of multiple-support excitation, the guidelines indicate that the phenomenon should be accounted for whenever
soil conditions along the bridge belong to different soil categories. The
guidelines also present:
• A stochastic model of the motion at the supports (Der Kiureghian,
1996) that can be used either to generate samples of correlated motions
to be used in time-history response analysis or in linear random vibration analysis;
• The multiple response spectrum method (Der Kiureghian and
Neuenhofer, 1992), which provides a solution for the random vibrations
problem of a system subjected to multiple inputs based on the use of the
corresponding input displacement response spectra;
• A simplified proposal for time-history analysis employing indepen46
The main objective of the project, which was the production of the draft
guidelines and their application manual, has been met. In this respect
the Research Line was successful, since the product has been delivered
and its quality is believed to be high.
Though it wasn’t explicitly included into the remit for the Line, it must
be noted that the research group initially intended to cover in the guidelines both structural concrete and masonry bridges. In spite of the
research carried out, however, this more ambitious goal could not be
Research on this front was essentially under the responsibility of the
unit of Genova. This unit has produced during the three years of the
project a considerable amount of high-quality research that has been
regularly documented in the annual as well as the final reports, and it
is also available in research reports from the unit uploaded on the project website. Quoting from the final report the issues dealt with by the
unit cover the following: “i) statistical characterization of the Italian
bridge population; ii) mechanical models for solid clay brickwork,
needed for detailed and simplified structural models; iii) in field testing
of masonry bridges, aiming at the identification of the main mechanical
properties of the materials and of the bridge as a whole; iv) laboratory
testing of brickwork prisms; v) reduced scale testing aiming at identifying the load carrying capacity and the collapse mechanisms of shallow
and deep arches taking into account the fundamental collaboration of
the so called “non structural elements”; vi) reduced scale testing aiming at identifying the dynamic properties of shallow and deep arches
RESEARCH - Seismic behavior
taking into account the collaboration of the so called “non structural
elements”; vii) Limit Analysis procedures for the analysis of masonry
bridges taking into account the contribution of all the bridge elements;
viii) retrofitting techniques for the bridge and its components.” As it
may be seen all of the investigated topics are of clear scientific interest,
though not specifically relevant to seismic assessment of bridges. This
is simply the unavoidable consequence of the international lack of fundamental knowledge on the seismic behaviour of masonry bridges.
tion of expected loss related to any given bridge. Looking now at the
problem of bridge protection from an higher perspective, the attention
should be directed at the bridges as components of road links forming
a transportation infrastructure. The seismic performance of the single
bridge would then be put in relation with the performance of all other
bridges to be able to estimate the overall decrease in functionality of the
whole infrastructure. In this respect the very challenging problem of
determining the loss in traffic capacity of a damaged bridge represents
an essential element.
The research carried out within Research Line 3 has included the stateof-the-art into a document usable for assessing the protection level of
bridges against a number of limit-states.
There are certainly several areas where improvement is possible and
desirable, and in particular these are:
• The non-linear static analysis for bridges of complex geometry;
• The ultimate strength and deformation capacity of structural members
such as those encountered in bridge structures (e.g. polygonal multicell. hollow-core cross-sections)
• The generation of ground motions for multiple-support excitation.
While generated motions are being progressively replaced with recorded ones for the analysis of buildings, their use appears unavoidable for
the analysis of bridges whenever different motions must be considered
at the supports. Currently available procedures are in need of considerable improvement.
• The vast literature on SSI needs to be acquired and digested by structural engineers to become a practical tool. This a crucial aspect in view
of the displacement-based framework of the guidelines and the corresponding need for more accurate evaluations of deformations.
The guidelines do not cover the seismic isolation technique. The reason
for this choice is that the design of seismic isolation does not vary
between new and existing bridges. Seismic isolation, however, will certainly see much further diffusion in the coming years, for new as well as
existing bridges, while isolation device technology continues to evolve
rapidly with the ensuing need of developing appropriate analysis and
design techniques. In this respect this can be regarded as an ongoing
research topic.
To the extent that solutions to the problem of assessing the protection of
a bridge against its ultimate state can be considered to be sufficiently
mature, the next important passage is that of being able of estimating
structural and monetary damage as a continuous discrete function of
seismic intensity. Achievement of this goal would allow for the estimaRESEARCH - Seismic behavior
- Casarotti C. (2005), “Adaptive pushover-based methods for seismic
assessment and design of bridge structures” PhD thesis ROSE School,
Pavia, Italy.
- CEN (2005), “Eurocode 8 Part 2: Seismic design of bridges”
European Committee for Standardization, Brussels, Belgium.
- CEN (2005) “Eurocode 8 Part 3: Assessment and retrofitting of existing structures” European Committee for Standardization, Brussels,
- Chopra A.K., Goel R.K. (2002), “A modal pushover analysis procedure for estimating seismic demands for buildings” Earthquake
Engineering & Structural Dynamics Vol 31(3), pp. 561-582.
- Der Kiureghian, A. (1996), “A coherency model for spatially varying
ground motions” Earthq. Eng. & Struct. Dyn. Vol. 25, pp. 99-111.
Der Kiureghian, A., Neuenhofer, A. (1992), “Response spectrum
method for multi-support seismic excitations” Earthq. Eng. & Struct.
Dyn., 21: 713-740
- DM2008 (2008), “Nuove norme tecniche per le costruzioni” Decreto
Ministeriale del Ministero delle Infrastrutture 14/1/2008.
- FHWA (1995), Seismic Retrofitting Manual for Highway Bridges,
Publ. FHWA-RD-94 052, Federal Highway Administration.
- FHWA-ATC (1983), “Retrofitting guidelines for Highway Bridges”
Report ATC-06-2, Applied Technology Council, Redwood City,
- FHWA-MCEER (2006), Seismic retrofitting manual for Highway
Structures. Part 1- Bridges.
- fib (2007) “Seismic bridge design and retrofit – structural solutions”
Bulletin 39, International Federation for Structural Concrete.
- Franchin P., Pinto P.E., Noto F (2007), “A nonlinear dynamic model
for seismic analysis of earth-retaining diaphragm-walls” Proc 4th Int.
Conf. Earthquake Geotech. Engng, Thessaloniki, Greece.
- Franchin P., Pinto P.E. (2007), “Analysis Of Diaphragm-Type Bridge
Abutments Before And After Seismic Upgrading” Proc 1st US-Italy
workshop on seismic design and assessment of bridges, Pavia, Italy.
- Gerolymos N., Gazetas G. (2006), “Winkler model for lateral response
of rigid caisson foundations in linear soil” Soil Dyn. & Earthq. Engng
Vol.26, pp. 347-361.
- Gerolymos N., Gazetas G. (2006), “Development of Winkler model for
static and dynamic response of caisson foundations with soil and interface nonlinearities” Soil Dyn. & Earthq. Engng Vol.26: 363-376.
- Isakovic T., Fischinger M. (2005), “Higher modes in simplified inelastic seismic analysis of single-column bent viaducts”, Structural
Engineering International.
- Kappos A.S., Paraskeva T.S., Sextos A.G. (2005), “Modal pushover
analysis as a means for the seismic assessment of bridge structures”
Proc. 4th European workshop on the Seismic behaviour of irregular and
complex structures, Thessaloniki, Greece (Paper 49).
- Lupoi A., Franchin P., Pinto P.E. (2007), “Further probing of the suitability of push-over analysis for the seismic assessment of bridge structures” Proc. of COMPDYN’07, Crete, Greece
- Makris N., Gazetas G. (1991), “Dynamic pile-soil-pile interaction.
Part I: Analysis of Axial Vibration” Earthq. Eng. & Struct. Dyn. Vol.20,
- Makris N., Gazetas G. (1992), “Dynamic pile-soil-pile interaction.
Part II: lateral and seismic response” Earthq. Eng. & Struct. Dyn.
Vol.21, pp 145-162.
- Monti G., Pinto P.E. (1998), “Effects of multi-support excitation on
isolated bridges” Tech. Rep. MCEER 98/0015 pp. 225-247.
- Novak M. (1974), “Dynamic stiffness and damping of piles” Canadian
Geotech. Jnl Vol.11: 574-598.
The use of fiber reinforced polymer (FRP) materials for the strengthening
of masonry and concrete structures, represents a valid alternative to traditional techniques. Indeed, many advantages are provided by using
FRPs: lightweight, good mechanical properties, corrosion-resistant, etc.
In Italy, the use of FRP materials for reducing the seismic vulnerability
of existing structures has been allowed for the first time through O.P.C.M.
3274 and more recently by the D.M. 14.01.2008, that refer to the Italian
National Research Council Design Guidelines (CNR-DT 200/2004) for
the external strengthening of existing structures with FRP materials.
These guidelines provide, within the framework of the Italian regulations, a document for the design and construction of externally bonded
FRP systems for the strengthening of existing structures. In particular,
several issues concerning the seismic rehabilitation of Reinforced
Concrete (RC) and masonry buildings have already been dealt but a further investigation is still required.
Within this context, the main aim of this research line has been the
experimental validation of design indications provided by the CNR-DT
200/2004 guidelines.
The main topics investigated in this research task can be summarized
as follows:
- the mechanical behaviour of FRP materials;
- the cyclic behaviour of RC elements strengthened by means of FRP;
- the mechanical and chemical anchorage devices for FRP systems;
- the ductility increasing of RC columns confined with FRP;
- the RC joint strengthening with FRP;
- the masonry strengthening with FRP;
- the historical structure strengthening with FRP;
- the quality control and monitoring of FRP applications;
- the innovative fibers (steel fabrics, natural fibers, FRP grids, etc.) and
matrices (organic and inorganic);
- the mechanical behaviour of concrete structures reinforced with fiber
reinforced polymer bars;
- innovative strengthening techniques (near-surface mounted (NSM)
technique, FRP prestressed systems).
The research activity has been performed through experimental tests
and theoretical studies mainly devoted to the development of simple
methods of analysis and design rules in order to improve the indications
RESEARCH - Seismic behavior
provided by CNR-DT200-2004.
FRP are ideal products for structural retrofitting and seismic upgrading.
Nonetheless the small knowledge on the durability of the system is one
of the main drawbacks to the use of FRP reinforcement in Civil
Engineering. In particular, structural adhesives usually represent the
weakest point of the reinforced system and their mechanical behaviour
and durability performance need to be investigated. The first problem
in using composite materials for structural reinforcement is the determination of their mechanical properties. The bond between FRP and
concrete is a very important issue because the debonding is a very brittle failure mechanism and must be avoided.
According to performance-based design or seismic evaluation of RC
buildings, it is crucial to provide a correct evaluation of the strength and
ductility capacity of the RC columns and beams as well as of beam-column joint. Experimental and direct observation of damages occurred
during recent earthquakes strongly highlighted this need.
The effectiveness of FRP systems for seismic vulnerability mitigation of
masonry structures is still in debate, despite it has moved a huge interest, becoming the outstanding system in the market for this type of
Indeed CNR DT 200/2004 has been the first guideline to provide
design criteria for the FRP seismic strengthening of masonry buildings.
However, the retrofit design of masonry structures is still not a completely solved problem. This is due to the fact that the masonry structure is load dependent and thus the FRP could be placed in an inactive
area of the resistant mechanism. Furthermore, masonry can activate a
large number of local mechanisms which interact with global behaviour
of masonry buildings. The non linear seismic assessment of FRP reinforced masonry structures is included also in the D.M. 2008 rule. The
non linear analysis requires the knowledge of the constitutive law of the
masonry material both in the unreinforced or strengthened situations.
In the recent years, the scientific research has been focused on the safeguard of historical buildings. Accordingly, CNR DT 200/2004 has been
published in order to provide design criteria for the use of FRP systems
for strengthening existing structures and to avoid their incorrect application. The Guidelines deal with different types of FRP applications to
masonry and reinforced concrete structures and take into account the
important phases of quality control and monitoring that should follow a
strengthening application. Several aspects affect the effectiveness of
FRP systems such as the surface preparation and FRP installation.
Moreover, once FRP strengthening intervention has been carried out,
monitoring by non-destructive or semi-destructive tests should be performed to ensure the quality and effectiveness of the strengthening system. It is worth noting that due to the increased number of composite
material applications and in order to get a better understanding of the
RESEARCH - Seismic behavior
interaction force between FRP materials and the masonry substrate,
experimental tests are needed. In this research line, semi-destructive
and non-destructive techniques have been also investigated for the
quality control and monitoring of FRP applications to masonry structures, according to CNR DT 200/2004 Guidelines.
In order to guarantee an optimal organization of the research, the
Research Units have been grouped into the following ten Tasks, each
one with a specific topic:
- Task 8.1: the mechanical characterization of FRP systems at fixed
environmental conditions under cyclic actions;
- Task 8.2: the delamination under cyclic actions and design of anchorage mechanical devices for FRP systems;
- Task 8.3: the confinement of RC and masonry columns subject to combined flexure;
- Task 8.4: the strengthening in flexure and in shear of RC structural
elements with FRP fabrics and near surface mounted (NSM) rods;
- Task 8.5: the beam-column and beam-foundation joint reinforcement
with FRP;
- Task 8.6: the design criteria for the seismic retrofit of RC and RCmasonry composite structures with FRP;
- Task 8.7: the design criteria for the seismic retrofit of masonry structures with FRP;
- Task 8.8: the strengthening of masonry structural elements with FRP
- Task 8.9: the strengthening of masonry vaulted elements with FRP
- Task 8.10: the quality control and monitoring of FRP applications to
existing masonry and RC structures.
4.1 Task 8.1: the mechanical characterization of FRP systems at fixed
environmental conditions under cyclic actions
The aim of the sub-task was to mechanically characterize FRP systems.
The study has been focused on durability and mechanical behaviour of
structural adhesives and FRPs, the mechanical characterization of FRP
bars and strips, the values of the safety factors proposed in CNR DT
200/2004 and the effects of elevated temperatures and freeze-thaw
cycling on FRP.
Durability and mechanical behaviour of structural adhesives and FRPs
Several tests to determine the mechanical properties of composite mate49
Fig. 1- a) Execution of the punch-tool test and specimen after collapse, b) execution of the torsion test and specimen after
Fig. 2- a) Anchor system for large diameter GFRP bars, b) numerical analysis.
rials and structural adhesives have been performed. Conforming to the
ASTM requirements, the glass transition temperature (ASTM D3418),
porosimetry and the coefficient of thermal expansion (ASTM D360)
were determined. Adhesives were also tested under tensile (ASTM
D360), compressive (ASTM D695) and flexural loading (ASTM D790).
Adhesive shear strength was determined by punch tool tests (ASTM
D732). Finally adhesive cylinder specimens were tested under pure torsion load.
Adhesive dumb-bell specimens were prepared for tensile testing and
then artificially aged in an environmental chamber in order to analyze
possible detrimental effects on the adhesive mechanical properties.
Exposition to deicing salts, freeze-thaw cycles and moisture may in fact
deteriorate the mechanical properties with consequences on the durability performances of strengthened structures. Tensile tests were performed conforming to the requirements of ASTM D360. In all the conditioning treatments, significant losses in adhesive stiffness and tensile
strength were measured. The stiffness and tensile strength reductions
after exposure to salt spray fog solution may be approximated by
straight parallel lines as described in the Arrhenius life-temperature
relationship. Fatigue tests on adhesive dumb-bell specimens were
finally performed to attain the fatigue failure curves for the adhesive
riences for testing steel ropes and prestressing steel tendons and the
shape of the resin head from test investigation. Numerical analyses
were also performed to investigate the effects of anchor parameters such
as cone slope angle, thickness of resin head and friction coefficient
between the anchor body and the resin head. Pull-out and beam tests
were also executed.
Experimental studies and numerical analyses were developed to define
practical tests for the characterization of FRPs and adhesives mechanical properties. The main aim of this action was to provide the
“Composites Kit Test - COKIT”; a practical tool for professionals and
engineers operating in the field of FRPs applications and dealing with
FRP materials for structural retrofitting and rehabilitation. The technical document “Istruzioni per la caratterizzazione ed il controllo di
accettazione di materiali fibrorinforzati per il rinforzo strutturale –
COKIT ” was thus published and could represent an annex of the CNR
DT 200/2004 Recommendations.
Refinement of the safety factors proposed in Design Recommendations
The environmental conversion factors provided in the guidelines of the
Italian National Research Council (DT200) were analyzed on the basis
of the results of artificially aged adhesive specimens tested under tension. Exposition to deicing salts, freeze-thaw cycles and moisture leads
Mechanical characterization of FRP bars and strips
Tensile and relaxation tests were performed on FRP bars with particular attention to the gripping system. Then, experimental tests and
numerical simulations were performed to develop simple, economical
and effective systems for the characterization of composite materials
and adhesives. In particular, an anchor system for tension testing of unidirectional fiber reinforced plastic (FRP) bars of large diameter was
developed. In the system suggested each end of the bar is embedded in
a conical polymeric head that fits a conical hole inside the anchoring
device. In the anchor system, the anchor body shape came from expe50
Fig. 3- Stiffness and tensile strength retention for the structural adhesive subject to freeze-thaw cycles of five hours each
between –18° and +4°C for a total duration of about 2 months (FT), to salt spray fog for one month or three months (SF)
and to one month humidity (HU).
RESEARCH - Seismic behavior
Effects of elevated temperatures and freeze-thaw cycling on FRP laminates behavior
Fig. 4- a) Steel-CFRP specimen b) Reduction in stiffness of retrofitted specimens during fatigue tests; c) S-N curve and
comparison between the fatigue resistance of the steel-CFRP bond for a stiffness reduction of 5% (blue circles) and of 15%
(red squares) and that of EC3 welded detail categories.
to the deterioration of the mechanical properties of composite materials
and in particular structural adhesives. On the basis of the experimental
results, the safety factors suggested in the CNR DT 200/2004 recommendations may be considered as appropriate, but in aggressive environments the use of a slightly lower conversion factor seems to be more
Tests were performed to refine the safety factor of FRP-steel systems:
the fatigue behaviour of steel structures retrofitted by using FRP materials was investigated, S–N curves were defined and the fatigue resistance of the steel-CFRP bond was compared to the one of welded detail
categories described in the Eurocode 3.
After performing pull-pull delamination tests on FRP-concrete specimens, cylinders were obtained from each concrete prism. Based on
Eurocode 2 compressive and splitting tests were carried out to determine the conditioning effects on concrete degradation. As a consequence of the environmental conditioning, concrete characteristic
strength is assumed to increase by 16% for salt spray fog conditioned
specimens and to decrease by 3% for specimens subject to freeze-thaw
The performances at elevated temperatures and/or at freeze-thaw
cycling exposure of structural members strengthened by using externally bonded FRP laminates are mainly related to two aspects: the bond
behaviour between FRP and the member substrate; the mechanical
properties of laminates themselves. The latter aspect has been very limited experimentally investigated; only few tests have been performed to
evaluate the residual tension strength of FRP coupons after exposure to
elevated temperatures or freeze-thaw cycling. Thus, experimental tension tests on carbon FRP (CFRP) laminates both under controlled temperature and relative humidity conditions or after freeze-thaw cycles
exposure have been carried out. In particular, due to reduced capacity
that commercially available resins have to transfer loads over fibres
around glass transition temperature, Tg, two new systems based on
epoxy resin have been formulated and characterized by dynamic
mechanical analysis (DMA). The main goal of the new formulated systems was to increase Tg, the elastic modulus in the rubber region of the
resin and to improve their performances under freeze-thaw cycles. Two
different approaches were investigated. First a new epoxy system
(namely neat epoxy) was formulated and cured at 60°C after an hour at
room temperature. Secondly, in order to improve the mechanical properties of epoxy matrix by curing at room temperature, a nanocomposite
system was obtained by direct dispersion of preformed nanodimensioned silica particles to the neat epoxy resin.
Fig. 6- Specimen geometry; Temperature and relative humidity exposure profiles; test setup.
Fig. 5- Statistical distribution of the coefficient kG for specimens subject to a) salt spray fog and b) freeze-thaw cycles.
RESEARCH - Seismic behavior
The experimental results point out that the developed formulations of
epoxy resins provide a significant increase of ultimate strength and
strain of CFRP coupons both at room and elevated temperatures with
respect to commercial systems, without significant change of the elastic
modulus. Negligible influence of a low number of freeze-thaw cycles
was observed on the mechanical properties of coupons independently of
matrices. Experimental outcomes strongly confirmed that the use of
matrices characterized by higher values of Tg and elastic modulus in the
rubber region with respect to those traditionally available on the mar51
4.2 Task 8.2: the delamination under cyclic actions and design of
anchorage mechanical devices for FRP systems
Different experimental set-ups can be found in the scientific literature
dealing with FRP-concrete bond tests and it has been observed that different test methodologies may give different values of the debonding
force. This task research intended to define a standard FRP-concrete
bond test to be used to evaluate the maximum transmissible force by an
FRP anchorage, to be included in the new version of the Italian code for
design of strengthening interventions with FRP.
single shear push-pull test. All the tests have been performed under
displacement control of the FRP free end. In order to evaluate the variability of the results when different set-ups are adopted, the coefficient
of variation (COV) for each set of homogeneous experimental tests has
been calculated. The scatter of the results is in general small (COV
about 10%), lower than that of the tension strength of the concrete, usually equal to 20-30%. For the plates, the scatter of the results is similar for the different Labs, whilst for the sheets the dispersion is usually
higher. The results obtained by Lab 3 are very stable in both cases and
close to the mean values. This study allowed to define a set of rules for
the standardization of bond tests to be used to evaluate the maximum
transmissible force by an FRP – concrete anchorage.
Experimental Round Robin test on FRP concrete bonding
Cyclic tests of FRP-concrete debonding under cyclic loadings
An extensive experimental campaign on FRP-concrete debonding has
been carried out by five different Italian Laboratories (University of
Bologna, University of Naples Federico II, University of Sannio,
Polytechnic of Milan and University of Calabria). The tests were devoted to the definition of a standard test procedure for the bond strength
evaluation. According to the Round Robin procedure, 50 concrete
prisms (same batch) strengthened with CFRP plates and sheets have
been prepared by the same operator and subject to bonds test in five
different Laboratories. The sets of homogeneous specimens have then
been subject to bond test by five laboratories of the University partners
using different test set-ups (Figure 7).
Many strengthened structures are subjected to fatigue loads (i.e. roads
and railways bridges) or to shorter but more intense cyclic actions as
seism: in this cases, the FRP-concrete interface is subject to cyclic
stress regimes which can lead to premature debonding of the FRP laminate from the concrete substrate and cause the FRP failure in most
cases, unless appropriate local measures are taken to prevent it.
In order to develop a more economical design for FRP-strengthened
structures, research line investigated the debonding phenomenon of the
FRP reinforcements (both plates and sheets) from the concrete substrate under cyclic actions. In particular, tests on little prismatic specimens have been performed by applying FRP plates and sheets on concrete prisms and testing them under both monotonic and cyclic actions
without inversion of sign.
The experimental results of Single Shear Test (SST) performed on CFRP
reinforcement applied on little prismatic concrete specimens and characterized by high bond length values (400mm) showed that:
- the influence of load-unload cycles up to 70% of Pmax,M was negligible for CFRP sheets and plates;
- a low number of load-unload cycles (40) up to 90% of Pmax,M reduced
the debonding load of about 10% in the case of CFRP plates but did
not affect particularly the bonding behaviour of CFRP sheets;
- by increasing the number of load-unload cycles (up to 300) between
70% and 90% of Pmax,M, the debonding load of concrete specimens
reinforced with CFRP sheets decreased by a percentage factor equal
to 10%;
- the transfer of shear stresses at the FRP-to-concrete interface due to
the CFRP reinforcement bond length larger than effective one,
allowed to mitigate noticeably the effect of cyclic actions imposed up
to 90% of Pmax,M;
- a degradation of interface behaviour has been recorded after the onset
ket, could allow to overcome one of the main limit of FRP laminates
related to their poor performances under elevated temperatures.
Twelve specimens (6 strengthened with sheets, 6 strengthened with
plates), with two different bonded lengths (100 mm and 400 mm), have
been tested by each laboratory, repeating three times the same type of
test. As for the test set-ups (Figure 7), all the Laboratories adopted a
Fig. 7- Experimental set-ups adopted by the five different Laboratories.
RESEARCH - Seismic behavior
of debonding, with reduction of maximum shear stress;
- the effects of cyclic actions were more significant on plates rather than
sheets and its influence increased with number of cycles;
Moreover similar SST tests performed on CFRP reinforcements characterized by lower bond length values (50-250mm) allowed to observe
- design relationships provided by Teng et Al. and by main international codes for evaluating the effective bond length values are conservative for both sheet and plate reinforcements;
- referring to plates the effective bond length values, experimentally
evaluated by means of the monotonic tests, were noticeably lower than
- even if reinforcement bond length values were significantly low, the
cyclic tests outcomes confirmed that the influence of load-unload
cycles up to 70% of Pmax,M was negligible for CFRP sheets and plates
due to elastic behaviour characterizing FRP-concrete interface up to
such load level;
- Also a further low number of load-unload cycles (10) up to 90% of
Pmax,M did not affect particularly the bonding capacity of CFRP reinforcement due to the transfer of shear stresses at the FRP-to-concrete
interface. Such transfer was more significant on plates rather than
sheets: nevertheless, bond lengths particularly conservative for plates
allowed to better mitigate cyclic action effects;
- in order to better predict design bond length values, using two different relationships for sheets and plates, respectively, could be worthwhile.
4.3 Task 8.3: the confinement of RC and masonry columns subject to
combined flexure
The main goal of this research task has been to validate the design
equations provided by CNR DT200 for the confinement of RC and
masonry members. In particular, experimental tests have been carried
out on both real scale and scaled columns wrapped by using traditional FRP (CFRP and GFRP) or an innovative typology of FRP system
made of basalt material fiber.
Confinement of real scale RC columns subject to axial load
An experimental campaign has been carried out on full scale reinforced
concrete (RC) columns concentrically loaded and confined by means of
FRP (Glass FRP and Basalt FRP). Five series of tests were planned, for
each series a reference unconfined column was tested and used as
benchmark. The test matrix has been designed to assess the confinement effectiveness: applying the same reinforcement ratio and checking the effect of the shape, the side aspect ratio and the area aspect
RESEARCH - Seismic behavior
Fig. 8- Debonding force for sheets with anchorage length (a) L=400 mm, (b) L=100 mm, and for plates with anchorage
length (c) L=400 mm, (d) L=100 mm.
Fig. 9- Coefficient of variation (COV) of the debonding force for (a) sheets and (b) plates.
ratio; applying different FRP reinforcement ratios and checking the
confinement sensitivity to the number of plies.
Tested specimens represent real scale building columns designed
according to dated codes for gravity loads only. The design concrete
strength is 23.1 MPa to simulate concrete mixes used in past decades.
Concrete cylinder specimens per each casting have been prepared in
order to characterize the concrete with standard procedures. The used
steel is characterized by a yield strength of 414 MPa and a modulus of
elasticity of 200 GPa. Particular care has been devoted to construction
details, namely: hooks, longitudinal and transverse steel reinforcement
ratios and concrete cover specifications. Special care has been taken to
avoid local failure at the top and the bottom ends of the columns placing steel ties with reduced spacing. Internal steel (bars and ties) reinforcement ratio is the same for each group of specimens, designed per
minimum code requirements. The minimum specimen dimensions are
360 x 510 mm2. The load has been applied concentrically under a displacement control rate. The load has been conducted in five cycles in
increments of one fifth of the expected capacity for each specimen.
Each loading-unloading cycle has been repeated once. Strain gages,
potentiometers and LVDTs are used for strain and displacement data
acquisition. In particular, strain gages have been applied on column
surface and internal bars whereas LVDTs have been placed in order to
obtain vertical and horizontal column displacement. Strain data acquisition have been obtained by strain gages applied on FRP sheets too.
Fig. 10- Test Matrix.
(a) Column S-1-5GA
(b) Column R-1-8H
Fig. 11- Columns strengthened by means of glass (a) and basalt (b) FRP.
The data have been elaborated in order to investigate on volumetric
strain and Poisson ratio as a function of load level. The main results of
the experimental campaign can be summarized as follows: a significant
increasing in the axial displacement and a little increase in the ultimate
load of the FRP strengthened columns compared to their benchmark.
The hollow columns have shown a failure mode characterized by
bulging, followed in such case by the rupture of the fibers. A more evident failure mode has been shown by the remaining columns for which
the fiber rupture has always accompanied the concrete spalling.
Confinement of RC cylindrical specimens strengthened by means of
basalt fibers and inorganic matrix
The effectiveness of such system as a confinement technique has been
analyzed by means of an experimental campaign on concrete cylindrical specimens. The effectiveness of the proposed technique is assessed
by comparing different confinement schemes: 1) uniaxial Glass Fibre
Reinforced Polymer laminates; 2) alkali-resistant fibreglass grid bonded with a cement based mortar; 3) bidirectional basalt laminates pre-
Fig. 12- Load versus vertical and horizontal strains.
RESEARCH - Seismic behavior
Fig. 13- Typical failure modes.
impregnated with epoxy resin or latex and then bonded with a cement
based mortar; 4) cement based mortar jacket. The main objectives of the
experimental program were: a) to investigate on the effectiveness of
confinement based on basalt fibres pre-impregnated in epoxy resin or
latex and then bonded with a cement based mortar (BRM); and b) to
compare the performance (in terms of peak strength and ultimate axial
strain gains) of different confinement techniques using advanced materials with respect to GFRP laminates jacketing.
The investigation was carried out on 23 concrete cylindrical specimens
with a diameter of D = 150 mm and a height of H = 300 mm.
Fig. 15- Failure modes.
schemes were experimentally analyzed in order to evaluate and compare the effectiveness of the proposed strengthening techniques: 1) uniaxial glass FRP laminates (GFRP) wrapping; 2) uniaxial carbon FRP
(CFRP) laminates wrapping; and 3) uniaxial basalt FRP (BFRP) laminates wrapping. In particular 9 tests, were performed on square tuff
masonry (external tuff blocks and inner core filled with tuff chips and
mortar) scaled columns (mass density equal to about 1530 kg/m3): side
average dimension equal to 220mm; and average height of about 500
mm corresponding to 8 courses of tuff bricks (height-width ratio equal
to 2.27). Masonry was made by scaled yellow Neapolitan tuff bricks
(50x50x100mm) and a pozzolan (local volcanic ash) based mortar
(thickness of 12mm). Further 9 tests were performed on square clay
brick masonry scaled columns (mass density equal to about 1700
kg/m3): side average dimension equal to 260 mm, and average height of
about 560 mm corresponding to 8 courses of clay bricks (height-width
ratio of 2.20). Masonry was made by clay bricks (55x115x255 mm) and
a pozzolan (local volcanic ash) based mortar (thickness of 13 mm).
Fig. 14- BRM wrapping installation procedure.
Experimental outcomes showed that:
• BRM confining system could provide a substantial gain both in compressive strength and ductility of concrete members inducing a failure
mode less brittle than that achieved in the GFRP wrapped members;
• lower performance were observed by concrete confinement provided
by a primed glass fiber grid bonded with cement based mortar with
respect to BRM and almost no influence was generated by the jacketing with mortar only.
• maximum ultimate axial strain increases were provided by GFRP
laminates wrapping.
Confinement of rectangular masonry columns subject to axial load
An experimental campaign dealing with 18 square cross-section both
listed faced tuff and clay brick masonry scaled columns subjected to
uniaxial compression load. In particular, three different confinement
RESEARCH - Seismic behavior
Fig. 16- Specimen details (dimensions in mm): (a) tuff masonry; (b) clay brick masonry.
Masonry columns were tested through monotonically applied axial compressive loading under displacements control mode with a rate of 0.005
The experimental outcomes showed that:
• GFRP and CFRP jackets led to similar compressive strength gains
on tuff masonry columns under axial loads.
• GFRP and BFRP confining system led to similar compressive
strength gains of brick masonry columns under axial loads. BFRP wrapping was more effective in terms of global ductility increase (i.e. ulti55
a significant increase in load carrying capacity and ductility after FRP
strengthening, which identified the columns as ductile elements despite
the brittle nature of the unconfined masonry. Differences in mechanical
behavior, due to the geometry of the columns, to the nature of different
materials, to different strengthening schemes, and to the amount of reinforcement, have been taken into account. The calibration of design
equations recently developed by Italian National Research Council,
CNR was conducted to compare analytical prediction and experimental
Fig. 17- Stress-axial strain relationships and specimens’ failure mode: (a),(c) and (e) tuff masonry; (b),(d) and (f) clay brick
mate strain gain equal to 413% and 259% for BFRP and GFRP wrapping, respectively) even if the mechanical external reinforcement ratio
of FRP laminates was lower then GFRP ones; such result could be
explained by the higher values of ratios efl/efu recorded on BFRP laminates.
• The use of high values of laminates unit height may significant reduce
the effectiveness of FRP wrapping systems since it could be detrimental to the quality of confinement execution.
• The presence of voids and protrusions on masonry members reduces
the ultimate transverse strain on FRP reinforcement with respect to that
typically achieved on concrete members.
Another experimental campaign has been carried out in order to show
the behavior of columns built with clay or with calcareous blocks, commonly found in southern Italy, especially in historical buildings.
Rectangular masonry columns were tested for a total of 33 specimens;
uniaxial compression tests were conducted on columns taking into
account the influence of several variables: different strengthening
schemes (internal and/or external confinement), curvature radius of the
corners, amount of fiber-reinforced polymer (FRP) reinforcement,
cross-section aspect ratio and material of masonry blocks. Materials
characterization was preliminarily carried out including a mechanical
test on plain masonry. For all cases the experimental results evidenced
Fig. 18- Limestone and clay brick masonry specimens.
The results obtained from the experimental campaign confirmed that
innovative strengthening techniques, using FRP sheets and bars, are
effective when confinement of masonry compressed elements is needed. Two types of masonry were investigated: the first made with clay
bricks, the second made with limestone blocks. Even if the properties
of the constituent materials were different, in both cases a significant
increase was measured in terms of peak load and ultimate axial deformation. Two construction schemes were considered: full core and hollow-core columns; the last type reproduces the patterns often found in
historical buildings. External and internal FRP confinement were tested, separately and combined. The proposed techniques are strongly
recommended when a seismic retrofit is needed, since the external confinement introduces a plastic behavior of the compressed masonry
which indicates a large capacity in storing elastic energy which is taken
by the fibers placed in the transverse direction. The presence of interRESEARCH - Seismic behavior
nal bars used as an internal confinement system is recommended in
addition to external FRP layers if ductility constitutes a main issue,
since in columns strengthened only with bars the ultimate load was
increased but brittle behavior of unconfined masonry remained.
Columns with hollow core also showed a significant increase of
mechanical properties when confinement was applied, especially in the
cases of GFRP external sheets combined with internal bars.
Confinement of circular masonry columns subject to axial load
An extended experimental investigation has been performed in order to
show the mechanical behavior of circular masonry columns built with
calcareous blocks that may be commonly found in Italy and all over
Europe in historical buildings. Different stacking schemes were used to
build the columns, aiming to simulate the most common situations in
existing masonry structures. Carbon FRP sheets were applied as external reinforcement; different amounts and different schemes of confining
reinforcement were studied. The experimental program included a new
reinforcement technique made by using injected FRP bars through the
columns cross section. The structural behavior of masonry columns
damaged under different levels of load and strengthened by using FRP
reinforcements has been also investigated.
Fig. 20- Specimens after failure.
• Displacement capacity resulted increased in all cases; strengthened
columns tested showed an extended postpeak plastic branch in the load
versus displacement curves;
• Columns confined with three 100 mm wide sheets showed higher
mechanical properties with respect to the same columns confined with
two 150 mm wide sheets;
• Damage caused by overloads applied in the precracking stage before
strengthening did not reduce the mechanical properties of FRP-confined columns;
• Presence of internal FRP rebars acted as an effective confining system for cross sections composed by four blocks;
• Application of design equations by Italian CNR furnished conservative results for complete FRP wrapping, whereas prediction of strength
for masonry confined with CFRP strips showed a reduced scatter with
respect to experimental results.
4.4 Task 8.4: the strengthening in flexure and in shear of RC structural
elements with FRP fabrics and near surface mounted (NSM) rods.
A new technique for the shear and flexural strengthening of RC structural elements has been investigated in this research task. In particular,
the use of Near Surface Mounted rods (NSM) for structural upgrading
has been deeply analysed by both analytical and experimental investigations.
Further, the effectiveness of FRP laminates, traditionally used to
strengthen RC or masonry members, has been investigated with reference to full scale prestressed concrete (PC) girders.
Fig. 19- Geometry and dimensions (mm) of columns.
Important remarks follow:
• High increase in ultimate strength and strain were evident after
• Complete FRP jacketing was much more effective than discontinuous
RESEARCH - Seismic behavior
NSM bars shear contribution: a calculation procedure
A calculation procedure suitable for practitioners has been developed
by simplifying a more sophisticated predictive model recently developed (Bianco 2008; Bianco et al. 2009a-b). That procedure briefly consists of: a) evaluating the average structural system composed of the
average-available-bond-length NSM strip confined to the corresponding concrete prism whose transversal dimensions are limited by the
spacing between adjacent strips and the beam cross section width
(Figure 20); b) determining the comprehensive constitutive law of the
average system above (Figure 21); c) determining the maximum effective capacity that the average system can attain during the loading
process of the strengthened RC beam by imposing a kinematic mechanism and d) determining the NSM shear strength contribution by summing the contribution provided by each strip. The constitutive law
(Figure 22) and in turn the equations to determine the maximum effective capacity assume different features depending on the main phenomenon characterizing the ultimate behaviour of the average structural system of the specific case at hand. Hereinafter, for the sake of brevity, the main features of that computational procedure are shown only for
the case of shallow concrete fracture (u = 4) and a resulting resisting
bond length whose value is equal to the effective bond length (Figure
23). Further details can be found elsewhere (Bianco 2008).
The predictions obtained by that calculation procedure were also
appraised on the basis of experimental results (e.g. Dias et al. 2007).
The maximum effective capacity for the case of shallow concrete fracture and a resulting resisting bond length whose value is equal (u = 4)
to the effective bond length can be evaluated by:
A Csf
sf 2
Vfi,eff = 1 A1C1 Ld f,max+ 2 2 arcsin(1–A3f,maxLd)+
+(1–A3f,maxLd) 1–(1–A3f,maxLd)2 –
L J l3sin(q+
A1= p 3
C1=1– 0 2 1 ;
A2=LpJ3l; A3=
C2sf =
Actual Vf and design value Vfd of the NSM shear strength contribution
can be obtained as follows:
Vfd= Vf = (2Nf,int
Vfi,eff sin
where Rd is the partial safety factor divisor of the capacity that can be
assumed equal to 1.1-1.2 according to the indeterminateness of the
input parameters.
Bond between NSM bars and surrounding concrete: experimental and
analytical investigation
Fig. 21- Main features of the calculation procedure: a) average-length NSM strip and concrete prism of influence, b) adopted
local bond stress slip relationship, c) NSM strip confined to the corresponding concrete prism of influence and semi-pyramidal
fracture surface, d) sections of the concrete prism.
Fig. 22- Possible comprehensive constitutive law of an NSM CFRP strip confined within a prism of concrete: (a) concrete
that reaches the free extremity (u=1) or strip tensile rupture (u=2), (b) superficial and/or absent concrete fracture and
ultimate resisting bond length smaller (u=3), equal (u=4) or larger (u=5) than the effective bond length and (c) deep concrete
fracture (u=6).
Fig. 23- Maximum effective capacity along the CDC for the case u=4: a) comprehensive constitutive law; b) capacity
Vfi,CDC (;); and c) imposed end slip Li,CDC (;); distribution along the CDC for different values of the CDC opening angle and d) effective capacity as function of the angle .
Pull-out test were carried out to investigate both the qualitative and
quantitative influence of some of the involved parameters on the bond
performance (De Lorenzis and Galati 2006, Galati and De Lorenzis
2006). Those parameters encompass: ratio between depth and width of
the slit, kind of epoxy-based adhesive used as binding agent, distance
of the NSM bar from the edge of the concrete prism, distance between
adjacent bars and employment of external FRP strips used to confine
the joint. Tests were carried out by means of a tangential-pull device to
apply the load, LVDT transducers to measure the slip at both the loaded
and unloaded extremity and strain-gauges throughout the adhered
length of the bar to measure the deformations along the joint. The measured quantities were processed to obtain the local bond stress-slip
relationship for the different values of the test parameters. Cyclic tests
were also carried out subjecting the joint at a limited number of cycles
whose maximum load was assumed equal to different percentages of the
peak static load. The cyclic tests were useful to evaluate the joint residual strength such as the one following a seismic action.
An analytical investigation has followed the experimental program
above (Rizzo and De Lorenzis 2007-2009b). In fact, the local bond
RESEARCH - Seismic behavior
stress-slip relationship obtained in the pull-out tests has been modelled
by suitable analytical functions whose unknowns were calibrated for the
different values of the test parameters. The local bond stress-slip relationship obtained by the cyclic tests was also modelled by analytical
functions. Then, the numerical solution of the governing differential
equation has allowed the peak pull-out load be determined as function
of the available bond length.
The pull-out tests were also simulated by a FE model, both in the Linear
and Non Linear range. The Linear FE model was adopted to evaluate
the bond-induced stresses on a plane transversal to the bar, evaluating
the maximum stresses for different values of the geometrical and
mechanical parameters of the joint and estimating so, local tangential
stress inducing the first-cracking in both resin and concrete. After that,
a Non Linear model was developed by modelling: a) the several materials according to the fracture mechanics and b) concrete/adhesive and
adhesive/FRP interfaces by employing interface elements.
Experimental and analytical investigation on the shear strengthening
contribution provided by NSM FRP bars on RC beams
Four points bending tests were carried out on RC beams strengthened
in shear by NSM FRP bars (De Lorenzis and Rizzo 2006, Rizzo and De
Lorenzis 2006-2009a). Those beams were designed in such a way that
the theoretical failure mode, for both the strengthened and un-strengthened beams was due to shear-tension. Parameters investigated were:
spacing, type and inclination of the NSM bars and the shear-span-todepth ratio. Some beams strengthened by NSM strips were also tested
in order to assess the relative effectiveness of the two techniques. The
system of FRPs was extensively equipped to measure the deformations
in the bars crossing the CDC. Tests have highlighted the possibility of
a global failure modes consisting in the detachment of the strengthened
cover from the underlying beam core. Such mechanisms had not been
pointed out by previous investigations.
Two models were developed to predict the NSM shear strength contribution: a) one more simplified and b) a more sophisticated one. The former was based on the Mörsch truss and the employment of a perfectly
plastic local bond stress-slip relationship. The latter takes into account
a more realistic local bond stress-slip relationship and the interaction
between existing steel stirrups and NSM bars. The different local bond
stress-slip relationships obtained in the former phase of the investigation were employed to carry out some comparison. From those comparisons it was possible to point out the great importance of the fracture
energy as opposed to the shape of the local bond stress-slip relationship. This phase of the investigation has led to the development of useful formulae for the evaluation of the NSM FRP shear strengthening
RESEARCH - Seismic behavior
contribution to a RC beam.
Experimental investigation on full-scale prestressed concrete beams
strengthened by means of CFRP
Every year, several prestressed concrete (PC) bridge girders are accidentally damaged by over-height vehicles or construction equipment
impact. Although complete replacement is sometimes deemed necessary, repair and rehabilitation can be far more economical, especially
when the time and the social cost of the method are drastically reduced.
The numerous advantages provided by the use of FRP laminates are
leading in a sharp increase on their use for bridge construction
strengthening. Experimental investigations were conducted in order to
validate such strengthening technique on PC damaged members and
accurately assess the upper limit of damage amount beyond which FRP
laminates are no longer adoptable as repair solution.
Starting from such purposes, an experimental campaign was conducted
on five full-scale (13.0m long, 1.05m high) PC double T-beams with a
reinforced concrete slab, designed according to ANAS (Italian
Transportation Institute) standard specifications. One beam was used as
control, and the other four were intentionally damaged in order to simulate a vehicle impact by removing the concrete cover and by cutting a
different percentage of tendons (17% on two specimens and 33% on the
remaining two). The repair, by using externally bonded carbon FRP
(CFRP) laminates, aimed at restoring the ultimate flexural capacity of
the member, taking particular attention to the laminates anchoring system. In particular, one test was performed on the control beam (referenced as S1), two tests were carried out on intentionally pre-damaged,
to simulate an over-height vehicle collision, beams (named S2 and S3,
respectively) and the remaining two on pre-damaged specimens
upgraded by using two and three plies of CFRP laminates anchored by
using U-wraps (named S4 and S5, respectively).
In Figure 24 and Figure 25 the test setup and experimental load deflection curves are reported.
The experimental study has shown that: 1) a loss of strands equal to
17% and 33% caused a flexural capacity decrease equal to 20% and
26%, respectively; 2) to restore the ultimate flexural capacity of the
undamaged PC specimen by using CFRP laminates it is necessary to
prevent fibers debonding; 3) U-wraps (width wf= 100mm spaced at
pf=150mm) were able to significantly delay debonding but if damaged
existing concrete is patched by cementitious mortar, a perfect bond has
to be guaranteed during the cross section restoration to prevent localized debonding of longitudinal reinforcement and thus fully exploit the
potential effective FRP strain increase; 4) CFRP laminates increased
Fig. 24-- Test setup.
Fig. 25- Experimental load deflection curves.
both stiffness and flexural moment capacity of PC damaged beams
(maximum moment recover equal to about 12% and 20% for specimens
with 17% and 33% of strands loss, respectively; 5) the strengthening
intervention led to weak failure mode with a global ductility loss.
The experimental outcomes qualify the application of FRP technique,
already adopted in several cases of impacted PC bridges, as an effective tool to restore the flexural capacity of PC girders; however the calibration of theoretical expressions for the computation of the design
FRP strain level considering the benefits provided by anchoring systems is strictly necessary.
4.5 Task 8.5: the beam-column and beam-foundation joint reinforcement
with FRP
According to performance-based design or seismic evaluation of RC
buildings, it is crucial to provide a reliable evaluation of the strength
and ductility capacity of the beam column joints: experimental and
direct observation of damages occurred during recent earthquakes
highlighted this.
This task focused on some aspects: namely, cracking of the joint panel,
longitudinal reinforcement bars slipping are deformability sources and
they could alter the capacity and interaction of beam and column members and the joint itself. F.E.M. modeling was adopted as an assessment
tool. The finite element code TNO DIANA 9.1 was adopted to simulate
and to analyze numerically some real beam column joint sub assemblages, characterized by nonlinear mechanical properties and geometrical detailing, smooth bars, structural deficiencies, as commonly found
in existing buildings. Such deficiencies were analyzed by means of
parametric analyses to evaluate the possibility to apply external
strengthening on members characterized by poor concrete quality, low
transverse reinforcement ratios, inadequate confinement due to lacking
stirrups (especially in external joints), low bond performance of smooth
and ribbed longitudinal reinforcement in columns and beams.
Numerical analyses evidenced typical failure modes, crack patterns,
influence of mechanical and geometrical properties on the behavior of
joints. Some numerical/experimental comparisons were made based on
significative tests available in scientific literature (for instance performed by Prof. Shiohara working group) or tested, during the RELUIS
Project, by UNIBAS R.U., allowing the numerical F.E.M. model to be
validated. The behavior of the joints controls the global seismic behavior of an entire structure and a building in particular, so that its assessment is a crucial task in the strengthening design. Based on such analyses, the validity of analytical models for unstrengthened joints, available in scientific literature, was checked. Such check was supported by
local information provided by the detailed, refined, F.E.M. analyses.
This numerical tool was a primary tool to understand the damage evolution and to assess the reliability of the main assumptions, equations
and procedures according to the “Quadruple flexural resistance in reinforced concrete beam-column joints” (Shiohara, 2001) proposed by
Prof. Shiohara working group. To provide a direct, practical tool, oriented to the profession more than a nonlinear refined F.E.M. analysis,
it was evaluated the opportunity to extend such consolidated model to
the case of externally bonded FRP strengthening of beam column joints.
This model is based on the solution of a system of equilibrium equations
referred to the four rigid bodies in which the joint can be ideally divided, sometimes neglecting compatibility. This model is able to account
for different failure modes and for the contribution of externally bonded FRP strengthening. The main assumptions can be recalled:
• Diagonal cracks in the joint form an angle of about 45°
• Normal concrete stresses acting on the main cracks can be reduced
to an equivalent force
• Longitudinal reinforcement provides only axial forces, so that any
dowel action is neglected
• There is a global symmetry both on the horizontal and vertical plane
The joint shear can be evaluated according to the sketch in Figure 26.
Vf = T + C's + C'c – Vc
Vf = T + T' – Vc
where T and T'are the tractions in the longitudinal bars at the joint section, respectively; C'c is the compression force in concrete, while C's is
the compression force in the bars. Vj is the joint shear; and Vc is the colRESEARCH - Seismic behavior
Fig. 26- Details of the forces due to the elements converging in the joint panel.
Fig. 28- The joint system and the external forces.
umn shear. In Figure 27 the joint division is shown: there are four rigid
and interacting bodies. Each body can be associated to three equilibrium equations. The symmetry of the joints allows the system to be
reduced to six equations. Moreover, Figure 28 shows the relation between the column shear, Vc, and joint shear, Vb, based on the equation
Vb = m Vc, where m = lc / lb.
umn length; and lb is the beam length.
(a) Forces in reinforcement and FRP
(b) Forces in concrete
Fig. 27- Internal forces.
The independent equilibrium equations are five. Horizontal and vertical forces equilibrium and moment equilibrium based on point o (the
middle of the joint panel) can be expressed as follows:
mV + j (F –F )+ j (F –F )– C =0
2 c 2b 1 2 2c 3 4 2 2
Where Nb is the horizontal force in the beam; F1, F2 are tractions in the
longitudinal bars of the beams at the joint section; F3, F4 are tractions
in the longitudinal bars of the columns at the joint section; F5 is the
traction in the stirrups spread along the height of the joint and reduced
to an equivalent force; F6 is the traction in the FRP reinforcement in
the x direction and reduced to an equivalent force; C1, C2 are compressions as shown in Figure 27b; jb, jc are the internal lever arms in the
beam and column respectively; q is the inclination angle of the main
cracks and assumed equal to 45°; Vc is the column shear; lc is the colRESEARCH - Seismic behavior
Horizontal and vertical equilibriums of forces follow:
Where Nc is the vertical force in the column; F7 is the traction in the
FRP reinforcement in the y direction and reduced to an equivalent
To evaluate the column shear capacity of an unstrengthened joint, the
F1, F2 and F5 forces can be assumed equal to the yielding forces of the
corresponding steel reinforcement while F6 and F7 are equal to zero.
The five unknowns are Vc , F3, F4 , C1 and C2 and can be evaluated solving the system (3)-(7).
To evaluate the column shear capacity of a strengthened joint, the F6
and F7 forces can be equal to the strength capacity of the external FRP
reinforcement both in the x and y direction, respectively. To account for
both the tensile failure of the strengthening or for a possible FRP
debonding, the FRP capacity is given by the minimum between tensile
strength and debonding force evaluated, for instance, according to CNR
4.6 Task 8.6: the design criteria for the seismic retrofit of RC and RCmasonry composite structures with FRP.
The applications of Carbon FRP (CFRP) and Glass FRP (GFRP) materials have grown during last years; at present, seismic applications have
become comparable if not more frequent than those related to lacks due
to gravity loads. The Italian guidelines for FRP interventions (CNR-DT
200, 2004), deal with the use of composite materials to seismically
upgrade under-designed RC and masonry structures. From seismic
standpoint, FRP strengthening is regarded as a selective intervention
technique. Based on the main deficiencies of the existing structure, the
driving principles of the intervention are based on two main strategies:
1) preventing potential brittle failure mechanisms (i.e. shear failure, lap
splice failure, buckling of longitudinal reinforcement in compression)
and “soft story” collapse mechanism, and 2) increasing global deformation capacity of the structure, either by enhancing the ductility of
plastic hinges without their relocalization or establishing a correct hierarchy of strength by relocalizing the plastic hinges according to capacity design criteria. The retrofit strategy is obtained by combining the
above principles; the definition of the retrofit scheme depends on the
gap between actual and target performance of the specific structure,
costs, functional characteristics and importance of the structure.
Experimental studies, aimed at validating the effectiveness of FRP to
achieve the above goals, are reported in the following.
Seismic strengthening of an under-designed RC structure with FRP
The outlined seismic strengthening strategy effectiveness was experimentally investigated within the European research project SPEAR
(Seismic PErformance Assessment and Rehabilitation). The structure
under examination was designed and built with the aim of creating a
structural prototype featuring all the main problems normally affecting
existing structures: plan irregularity, dimensions of structural elements
and reinforcement designed by considering only gravity loads, smooth
reinforcement bars, poor local detailing, insufficient confinement in the
structural elements and weak beam column joints (see Figure 29). The
structure was subjected to pseudo-dynamic tests, both in its original
configuration and retrofitted by using GFRP.
tively. The design of the rehabilitation was based on deficiencies underlined by both the test on the ‘as-built’ structure and the theoretical
results provided by the post-test assessment. They indicate that a retrofit intervention was necessary in order to increase the structural seismic
capacity; in particular, the theoretical results showed that the target
design PGA level of 0.30g could have been sustained by the structure
if its displacement capacity was increased by a factor of 48%. In order
to pursue this objective, the retrofit design strategy focused on two main
aspects. First it was decided to increase the global deformation capacity of the structure and thus its dissipating global performance; such
objective was pursued by confining column ends with two plies of
GFRP laminates. Moreover, the second design key aspect was to allow
the structure to fully exploit the increased deformation capacity by
avoiding brittle collapse modes. To achieve this goal corner beam column joint panels were strengthened by using two plies of quadri-axial
GFRP laminates as well as a wall-type column for its entire length with
two plies of the same quadri-axial GFRP laminates used for the above
joints (see Figure 30).
Fig. 30- Column confinement and shear strength of corner joints (a); shear strength of wall-type column and retrofitted
structure overview (b).
Fig. 29- (a) Plan view and (b) 3D view of the SPEAR structure.
The structure in its original configuration was subjected to experimental tests with maximum peak ground acceleration (PGA) of 0.20 g. Since
both theoretical and experimental results showed that the ‘as-built’
structure was unable to withstand a larger seismic action, a retrofit
intervention by using FRP laminates was designed. Once the design of
the GFRP retrofit was provided, the structure was subjected to a new
series of two tests with the same input accelerogram selected for the ‘as
built’ specimen but scaled to a PGA value of 0.20g and 0.30g, respec62
Fig. 31- Base shear – top displacement curves for ‘as-built’ and FRP retrofitted structure.
The assessment of structural global performance, before and after the
strengthening intervention, was performed by nonlinear static pushover
analysis in longitudinal direction (positive and negative X-direction,
PX and NX, respectively) and in transverse direction (positive and negative Y-direction, PY and NY).
In Figure 31, the theoretical base shear-top displacement curves for the
RESEARCH - Seismic behavior
Fig. 33- Details of the designed rehabilitation system.
Fig. 32- – Theoretical seismic performance comparison at 0.3g PGA between ‘as-built’ and FRP retrofitted structure.
‘as built’ and FRP retrofitted structure are depicted with reference to
direction NX (where the maximum capacity-demand gap was recorded
for the ‘as-built’ structure at the significant damage limit state LSSD).
Figure 31, clearly shows that the FRP retrofit is able to greatly increase
the global deformation capacity of the structure, slightly affecting its
strength. The comparison between the seismic structural capacity and
both elastic and inelastic demand is reported in Figure 32 (for direction
NX) by using the Capacity Spectrum Approach (CSA) (Fajfar, 2000).
Figure 31 clearly shows that column confinement provides the structure
with significantly enhanced ductility, allowing it to achieve the theoretical inelastic demand by only modifying the plastic branch of the capacity curve.
After that columns and joints were wrapped with GFRP, the retrofitted
structure was able to withstand the higher (0.30g PGA) level of excitation without exhibiting significant damage. After tests, FRP was
removed and it was shown that the RC core was neither cracked nor
damaged. The comparison between the experimental results provided
by the structure in the ‘as built’ and GFRP retrofitted configurations
highlighted the effectiveness of the FRP technique in improving global
performance of under-designed RC structures in terms of ductility and
energy dissipation capacity.
Seismic rehabilitation of PC bridges by using FRP and SRP materials
An innovative strengthening technique based on the combined use of
Fiber Reinforced Polymer (FRPs) and Steel Reinforced Polymer (SRPs)
RESEARCH - Seismic behavior
has been investigated with reference to a PC bridge, named “Torrente
Casale” which is part of the Salerno-Reggio Calabria motorway. The
bridge, built in the seventies, has been recently enlarged (2001) in
order to satisfy the new traffic demands. Due to the recent issuing of a
new seismic Italian code, it was decided to assess the bridge capacity,
both for gravity and seismic loads, in relation with the new design provisions.
The bridge existing documentation has been investigated and both
destructive and non-destructive tests have been performed in order to
determine concrete and steel reinforcement mechanical properties.
Once the bridge geometry and the material mechanical properties were
determined, a theoretical analysis was performed showing that the
bridge piers (circular cross-section, diameter D=1m, and total height
H=6m) were not adequate to sustain the seismic actions. Thus use of
SRPs spikes as columns’ flexural reinforcement combined with CFRP
laminates wrapping of columns ends has been investigated to increase
both member strength and ductility; the structural upgrade was completed by increasing the shear capacity of column cap through CFRP
U-wraps (Figure 33). The effectiveness of such technique with respect
to a traditional one, based on RC jacketing, has been assessed. The
main construction phases of the rehabilitation intervention are reported
in Figure 33.
4.7 Task 8.7: the design criteria for the seismic retrofit of masonry structures with FRP.
The main design problems concerning the seismic vulnerability mitigation of masonry structures are the strengthening of masonry panels for
in plane and out of plane actions, the improvement of FRP action by
means of clamping and connection mechanical devices and procedures
for the definition of the reinforcement layout in seismic analyses.
In order to test the design methods proposed by CNR DT200/2004, sev63
eral seismic studies of complex monumental masonry buildings have
been completed by carrying out detailed linear dynamic seismic analyses and FRP retrofit design. In the design several combinations of
strengthening techniques were explored in order to point out the real
applicability of the cited guidelines.
The obtained results cover the following areas:
a) Verification of the feasibility of a FRP design through simple modifications of the normal design activity;
b) Definition of the feasibility bounds for this retrofit technique in case
of monumental buildings;
c) Detail types and level of detailing required in real applications;
d) Open FRP design problems not covered by the guidelines or lacking
of necessary information;
e) Comparison of different modelling techniques making use of plate,
shell, beam, truss elements to represent the masonry structure;
f) Comparison of different material constitutive assumptions for the non
linear analysis of strengthened walls and buildings.
The main results of the completed research are summarized in the following:
a) The FRP reinforcement net design is easily implemented by a simple modifications of the normal design activity. The only needs are: integration of the stress distributions in order to obtain the stress resultants
and preparation of a verification sheet including the design rules for
bending and shear of a FRP reinforced masonry panel. Actually many
companies have implemented DT 200/2004 rules in freeware software
which can be helpful in this activity.
b) The feasibility bounds for this retrofit technique checked for three
monumental buildings analysed, are very different if stuck or mechanically fastened reinforcement is used. In fact, only slight diffuse
increase of resistance is obtained by using externally bounded FRP
nets. Owing to increase significantly the safety of the building, mechanical devices are mandatory. In this last case true building retrofit is possible.
c) The set of the detail types needed for practical applications is very
large and many connection types are not yet fully investigated although
practically employed. Fibre ropes, bars, fasteners, metallic inserts are
mixed in a way almost never covered by existing experiments and
d) Open FRP design problems include cement matrices, thermal
cycling, delamination in compression, real bond of the regularization
primer to rough masonry surfaces, behaviour of the mechanically fastened FRP elements after internal delamination.
e) Several modelling techniques have been employed in order to carry
out the seismic studies. Plate elements in linear dynamic analysis allow
for both in plane and out of plane evaluation, but non linear static
analysis with this type of discretisation is not a viable solution. Studies
on discrete representation of masonry walls by using a refined truss
structure to represent the compressive load paths inside the masonry
panels showed that this type of discretisation is very simple, allows for
complicated constitutive laws, allows easily equilibrium checks, and
produces very reliable load – displacement curves.
f) The basic choice of associate Mohr – Coulomb or Drucker – Prager
elastic plastic constitutive laws can be effective only in limited cases,
where dilatancy is not dominant. More effective analyses require a non
associate zero dilatancy rule, which is however not common in the professional engineering software. More refined damage rules are actually
under way, but not for normal design activity. Truss elements however,
allow to introduce complex behaviour by means of geometric discretisation, and by this way, crack tracking can be pursued, even if in a very
rough representation.
4.8 Task 8.8: the strengthening of masonry structural elements with FRP
The research activity has been performed through experimental tests
and theoretical study. In particular, the experimental program has been
carried out with reference to masonry panels strengthened with FRP
and with SRG and subjected to in plane loads, while the theoretical
study concerns the modelling of masonry walls strengthened with FRP.
Masonry panels and building strengthened with FRP
Several experimental tests have been performed with reference to
masonry panels strengthened with different FRP materials (carbon and
glass) arranged according to various configurations.
In-plane shear–compression tests have been performed on full-scale
tuff masonry panels consisted of two-layered walls with the inner part
filled with mortar and chips from yellow tuff blocks, considering different FRP materials and strengthening configurations. In particular, two
sets of panels have been strengthened with grid pattern carbon fibre
unidirectional strips (CFRP), made with three horizontal and vertical
strips on each face and two sets of panels have been strengthened with
the same layout, but doubling the number of plies. Similarly, four sets
of panels have been symmetrically strengthened with a grid pattern on
both sides of the panels, but with glass bidirectional fibre strips
(GFRP). A further set of panels has been made selecting a different geometric configuration by arranging FRP laminates along the diagonals of
both sides of the panels and considering both a CFRP and a GFRP
cross layout with either one or two FRP plies.
Important experimental evidences have emerged from the performed
tests underlying the role of the configuration of the FRP strengthening
RESEARCH - Seismic behavior
system on the failure mechanism of the tested panels. In fact, while in
some cases it has been observed that the debonding of the FRP strips
has been the main responsible of the panels failure, in few cases the
tensile rupture of the FRP strips has occurred. The different observed
failure mechanisms have particularly influenced the behaviour of the
strengthened panels both in terms of strength increase and post-peak
behaviour (fracture energy).
Further tests have concerned square masonry panels composed of clay
bricks, strengthened with different FRP configurations and subjected to
compression diagonal load. In particular, some of the panels have been
strengthened considering both vertical and horizontal CRFP and GFRP
strips, while other panels have been strengthened arranging FRP strips
along diagonal directions of the panels. Different levels of the strength
increase and different failure mechanisms have been observed. In particular, while localized cracks pattern have characterized both the case
of un-strengthened panels and the case of panels strengthened by FRP
only on one side, a more diffuse crack distribution has been observed
in the case of panels strengthened on the two opposite sides. In several
cases the debonding of the FRP strips has been also observed.
From the experimental studies conducted on FRP strengthened masonry panels, considerations useful for improving the Document CNRDT200/2004 have been deduced. In particular, it has been observed
that for the evaluation of the design shear strength of FRP strengthened
masonry panels (eqn. 5.16 of the CNR-DT200/2006), the contribution
of the masonry component can be evaluated through the eqn. 5.17 (used
in the case of reinforced masonry elements) only if the FRP shear
strengthening is coupled with FRP vertical elements fixed both at the
base and at the top of the panel. For this reason it has been suggested
to include in the CNR-DT200 specific design indications concerning
this aspect. In fact, it is important to underline that, when the FRP
shear strengthening system is not coupled with a flexural FRP strengthening system with efficient anchor elements, the contribution of the
masonry material in terms of shear strength is the same of the case of
un-reinforced masonry elements.
A further aspect deduced with reference to tests on masonry panels
strengthened by FRP has concerned the case of fibres arranged along
to the diagonal directions of the panel. From the experimental evidences and from global analyses it has been possible to affirm that the
contribution of the FRP can be evaluated only considering the component parallel to the shear load.
A scale model of a typical tuff masonry buildings has been constructed
and tested on the seismic simulation shaking table at the structural laboratory of CESI, Bergamo, Italy. In particular, the test procedure has
consisted in three phases: in the first phase dynamic tests have been
conducted on the un-strengthened prototype in order to induce some
RESEARCH - Seismic behavior
damages; in the second phase the prototype has been repaired by
selecting GFRP strips arranged according to the Italian Code
(Ordinanza 3431/2005, CNR DT 200/2004); in the third phase, dynamic tests have been performed on the strengthened prototype.
Experimental evidences have shown that the applied GFRP strips allow
to perform fast repair interventions in order to make operational masonry structures severely damaged by earthquake and at risk of aftershocks. In fact, the repaired prototype has showed reduced openings at
the horizontal joints (about one-third of the maximum opening observed
in the case of un-strengthened prototype) and an increasing of the lateral strength (about +34%). A further aspect emerged from this tests
has concerned the importance to provide effective anchorages for FRP
strengthening elements in order to avoid the delamination phenomenon
which is particularly influenced by cyclic actions.
Masonry panels strengthened with SRG
Beside the “traditional” Fiber Reinforced Polymers (FRP), the research
investigated the possibility of application of innovative composite materials, called Steel Reinforced Grout (SRG), based on high strength steel
wires (Ultra High Tensile Strength Steel) forming that are assembled
into a fabric and embedded within a cementitious grout. This application in fact could combine, to the traditional advantages proper of FRP,
the performances of this new material, reducing installation and material costs, and inducing an increase of ductility. Both composites, FRP
and SRG, can be used with perforated or solid brick to form a new
strengthening system, called LATLAM ring – beam, that can be used to
effectively construct the roof ring beams of a masonry structure. This
new system, can be also subjected to a pretension force. The following
conclusions may be drawn from the developed research:
- the analytical model developed to determine the mechanical behaviour of LATLAM ring – beams has shown good agreement with experimental results and can be incorporated in design provisions;
- the experimental tests, performed on full – scale prototypes of LATLAM ring – beams, demonstrated good results in terms of load carrying
- LATLAM ring – beams proved to be a good substitute, either under a
technical and economical perspective, of “traditional” reinforced concrete ring – beams.
Modelling and analysis of masonry walls strengthened with FRP
A further subject of the task research activity has concerned the modelling and the analysis of masonry elements strengthened with FRP.
Indeed, recent codes extended the use of displacement-based design
methodologies, such as the pushover analysis, to the case of masonry
buildings. Thus, different modelling approaches able to capture the
structural behaviour of masonry panels strengthened with FRP have
been examined: a macro-micro modelling approach; a macro modelling
approach; a frame model approach.
The first model relies on a homogenization approach combined with
limit analysis suitable for the evaluation of the collapse loads and failure mechanisms of FRP reinforced masonry panels. The application of
FRP strips on masonry has been treated adopting a simplified multi
step approach. In the first step the un-reinforced masonry, regarded as
a periodic heterogeneous material, has been substituted with a homogeneous macroscopic material using a homogenization technique. In
particular, an estimation of the homogenized unreinforced masonry
strength domain has been obtained by means of a micro mechanical
model based on the lower bound theorem of limit analysis. In the second step, FRP strengthening has been introduced on the already
homogenized masonry material.
The second model is based on the use of both 2D or 3D nonlinear
behaviour finite elements and interface elements. In particular, special
yield criteria coupled with nonlinear constitutive laws characterized by
softening response have been selected in order to simulate the behaviour of masonry material both in tension and compression. A special
nonlinear constitutive law has been also considered for the interface
elements in order to simulate the debonding mechanism of FRP which
characterizes in several cases the failure mechanism of masonry elements strengthened with FRP.
Finally, the third modelling approach consists of a simple but effective
1D frame element able to predict the response of masonry structures,
eventually reinforced with FRP materials. In fact, the proposed elements presents some peculiarities both for converting the geometry of a
masonry panel in the geometry of the equivalent frame and for accounting the nonlinear behaviour of masonry and FRP materials. Special
components have been also provided in the developed model in order
to account for the shear failure modes.
Several numerical applications have been carried out using the proposed modelling approaches and considering experimental cases
deduced from literature. The obtained results, regarding both unstrengthened and FRP-strengthened simple panels and masonry
façades, have shown the reliability of the proposed models to reproduce
the experimental behaviour of masonry elements underlying some
peculiarities due to the presence of the FRP strengthening system. In
fact, the micro-macro and the macro modelling approaches have particularly underlined the role of the FRP strengthening system to change
the stress path and, consequently, to induce different damage states
characterized by a diffuse crack path along the strengthened elements.
This aspect has revealed the important role of the FRP system to
increase the energy dissipation capacity of masonry structures. The
increase both in terms of strength and ultimate deformation capacity of
strengthened elements has been confirmed by all the modelling
approaches evidencing the capability of the simple frame model to capture the global response of masonry elements strengthened with FRP
and consequently the possibility of using this simple model in a design
process of the FRP strengthening system for masonry structures.
4.9 Task 8.9: the strengthening of masonry vaulted with FRP systems.
The aim of the research task is the development of models and procedures for the analysis of vaulted structures reinforced with FRP. In particular the behaviour of arches, vaults and domes subjected to seismic
action is studied in order to investigate the behaviour of these typologies as built and strengthened with FRP materials.
The study is performed with simplified analytical procedure and with
numerical methods. Experimental tests are further expected for the validation of the models.
The research activity has been focused on the experimental tests on the
topic available in literature for a comparison with the analytical and
numerical results.
On the basis of the acquired data, a simplified analytical model for the
evaluation of the ultimate load of arches and portal frames reinforced at
the intrados and/or extrados has been developed.
The main results are related to the definition of a methodology for the
evaluation of the ultimate load of one-dimensional vaulted structures
reinforced with FRP and subjected to vertical and horizontal forces.
Starting point of the procedure is the assessment that the FRP presence
does not allow the formation of the collapse mechanisms, which characterise the ultimate behaviour of the un-reinforced structure. As a consequence the cinematic approach of the limit analysis, usually adopted
for this kind of structure cannot be applied. The proposed analytical
model, starting from the analysis of the ultimate behaviour of the unreinforced structure, identifies the location of the hinges, up to make the
structure statically determined, and then to be solved with equilibrium
condition. Obviously in this case and contrarily to the limit analysis, the
mechanical characteristics of the masonry and of the FRP and mainly
the debonding phenomenon of the composite material, significantly
influence the structural response.
The model has been developed and applied to assigned geometries of
arches and portal frames, subjected to static vertical and horizontal
loads, with FRP reinforcement at the intrados. The obtained results
have been expressed in terms of interaction domains in the plane of the
vertical and horizontal forces.
Afterwards, the procedures for the evaluation of the ultimate behaviour
RESEARCH - Seismic behavior
of masonry arches reinforced with FRP have been extended to other
typologies, such as the masonry portal frames reinforced at the intrados
or extrados.
In particular has been defined the formulation of a secant linear relationship defined by the maximum load and by the related displacement
at the brittle failure, to be adopted for the estimate of the seismic behaviour of the structure. This model, even if simplified, can be suitable for
simulating the response of vaulted structures subjected to vertical and
horizontal loads, strengthened with FRP sheets. In this case, indeed, if
the FRP is applied at the intrados, the failure is due, generally, to its
delamination, while if the composite material is glued at the extrados,
the failure is due to compression of shear crisis of the masonry. Anyway
the global behaviour of the structure is often characterised by an almost
linear response up to the maximum load, ad generally with a brittle failure.
Numerical FEM procedures in bidimensional field, have been further
developed. The comparison between the results obtained with the two
methodologies has allowed the validation of the simplified models.
Parallel to the analytical-numerical models, an experimental program
has been set up, on vaulted structures reinforced with FRP.
Three circular arches, subjected to a key load, have been tested. The
first one represents the reference un-reinforced, the other two have
been strengthened with FRP at the intrados. The tests have been performed in displacement control, with the aim of evaluating the postpeak behaviour and the softening branches.
The results obtained have allowed the preliminary validation of the
model related to arches reinforced at the intrados.
The last phase of the research was devoted to the validation of the
developed models, through experimental tests on masonry vaulted elements. In particular, besides the preliminary tests in masonry arches
reinforced with FRP at the intrados, a test has been performed on a
masonry portal frame, in full scale, reinforced with FRP and subjected
to vertical and horizontal loads. The obtained results allowed the validation of the analytical models and assessed the validity of the assumed
simplified hypotheses.
Fig. 34- Portal frame under construction.
RESEARCH - Seismic behavior
Fig. 35- Portal frame during the test.
In the Laboratory of Structures and Materials of the Department of Civil
Engineering of the University of Rome “Tor Vergata”, has been realized
a masonry portal frame, constituted by two columns with rectangular
section (24cm x 49cm) and height of 2m, and an arch with internal
radius Ri = 130cm and square section with side equal to 24 cm.
The portal frame was constructed with masonry block on mortar beds,
and then reinforced with a layer of FRP at the intrados of the arch and
on the internal surface of the columns. No anchorage of the composite
has been given at the foundation level.
A constant vertical load (v) has been applied on the arch key and two
horizontal increasing loads, with equal intensity (H/2) are given at the
arch abutments, which simulate a seismic action.
The experimental results on both the arches and a portal frame appear
to validate the simplified models and the procedures developed in this
research program for the design and check of masonry vaulted systems
reinforced with FRP.
4.10 Task 8.10: the quality control and monitoring of FRP applications
to existing masonry and RC structures.
Several masonry and reinforced concrete structures have been utilized
for the application of fiber reinforced composite materials, with the aim
of carrying out quality control and monitoring tests, in accordance to
CNR DT 200/2004. For each structure, special working sheets have
been developed for a proper characterization of the building from a geometrical, mechanical and logistic point of view. On these structures
semi-destructive tests, non-destructive tests and delamination tests
have been performed.
Semi-destructive tests
This test consisting in pull-off and shear tearing tests conducted on different types of FRP materials, applied on the in-situ structures.
Moreover, pull-off tests and shear tearing tests have been conducted in
the laboratory of the Department of Structural Engineering of University
of Calabria on concrete elements as well, reinforced by carbon fiber
sheets and laminates. The latter tests results have been utilized for a
comparison with a number of experimental tests conducted at the
Universities of Bologna, Naples, Sannio and Milan, in the framework of
Task 8.2 Round-Robin tests, with the aim of obtaining a standard test
procedure for delamination tests and evaluating the scattering between
experimental results derived from tests conducted on specimens realized by the same worker, but tested in different laboratories. The tests
have been conducted using the same device used for outdoor tests. Pulloff tests, used to assess the properties of the strengthened substrate,
have been carried out attaching a thick circular 75 mm diameter steel
plate to the FRP and isolating it from the surrounding FRP with a core
drill, taking particular care in avoiding heating of the FRP system while
a 1-2 mm incision of masonry substrate was achieved. The test consists
in pulling off the steel plate by means of an ad hoc device (Figure 36a), obtaining the ultimate pull-off strength value expressed in kN
(Figure 36-c). Whereas, shear tearing tests are used to assess the quality of bond between FRP and masonry substrate. These tests can be
conducted only when it is possible to pull a portion of the FRP system
in its plane located close to an edge detached from the masonry substrate. The tests have been carried out using the same ad hoc device
used for pull-off tests. In particular, metallic elements have been set up
onto the masonry wall and through the FRP strip, with the aim of connecting the entire test device. Then, the FRP element has been tightened until collapse (Figure 36-b), obtaining the failure tearing force,
expressed in kN (Figure 36-d). For what concerns in situ tests, 16 reinforced concrete structures and 17 brick and stone masonry building
have been considered, for a total number of more than 300 tests. The
FRP materials have been applied in the form of strips having the
dimensions of 500x200 mm and 50x200 mm in the case of r.c. structures and the dimensions of 500x300 mm e 50x300 mm in the case of
masonry structures for the execution of pull-off and shear tearing test,
respectively. Both carbon, glass and natural fiber composites have been
According to CNR DT 200/2004, FRP application may be considered
acceptable if at least 80% of the tests return a pull-off stress not less
than 10% of masonry support compressive strength, or not less than
0.9-1.2 MPa in the case of reinforced concrete structures, provided that
failure occurs in the support itself. For what concerns shear tearing
tests, FRP application may be considered acceptable if at least 80% of
the tests return a peak tearing force not less than 5% of masonry support compressive strength, whereas it has to be not less than 24 kN in
the case of reinforced concrete structures.
For what concerns masonry structures, taking into account the compressive strength of the support, both pull-off and shear tearing experimental results respect the limit values suggested in CNR DT 200/2004.
In the case of concrete structures, pull-off results are in accordance to
CNR DT 200/2004, whereas in shear tearing test results the limit value
of 24 kN suggested for reinforced concrete structures, has never been
reached. This is probably due to the small dimensions of the FRP
strips. In fact, CNR DT 200/2004 Guidelines don’t provide any instruction about the FRP dimensions that should be utilized to reach the
above mentioned limit value.
In the case of pull-off tests, failure has always occurred in the substrate,
for each type of FRP material applied, as expected.
The semi-destructive tests conducted in the laboratory on concrete ele68
Fig. 36- Pull-off test (a) and shear tearing test (b); Pull-off test result (c) and shear tearing test result (d).
ments reinforced by CFRP sheets and laminates have been realized.
From mechanical characterization tests, a value of 25 N/mm2 for concrete compressive strength was found. The Research Unit of University
of Calabria tested 3 prisms reinforced by 3 CFRP strips and CFRP laminates applied on their surfaces. Pull-off and shear tearing tests have
been conducted, with the aim of studying the failure mode and debonding of FRP from the substrate.
The results of pull-off tests conducted on both strips and laminates
respected the limit values suggested in CNR DT 200/2004, and also in
these cases failure occurred in the substrate, showing the effectiveness
of the FRP application. On the other side, shear tests showed a different behaviour of the composite materials for strips and laminates. In
particular, in the case of laminates, the collapse occurred suddenly and
the whole composite debonded from concrete, with an ultimate value of
the shear force higher than 24 kN (limit value suggested in CNR DT
200/2004). In the case of CFRP strips, a partial delamination of the
composite occurred and the limit value was never reached.
Non-destructive tests
Non-destructive tests, consisting in thermographic tests were conducted on both brick masonry elements built in the laboratory and on the
same real masonry structures used for the above mentioned semidestructive tests, reinforced with different FRP materials. These tests
are usually carried out to characterize the uniformity of FRP application. Both, the active and the passive thermography technique have
been adopted, in which thermal energy is applied externally onto the
test object or the natural infra-red radiation emitted by the object due
to a sufficient exposure to sun light can be utilized, respectively.
RESEARCH - Seismic behavior
The first Infra-Red thermographic tests have been conducted in the laboratory of the Department of Structural Engineering of University of
Calabria on brick masonry macro-elements, which have been reinforced by carbon fiber strips placed in different directions onto the
specimen surfaces. From the test a mortar joint was clearly visible due
to the non plane substrate surface.
The in situ masonry and reinforced concrete structures have been utilized for thermographic tests as well, to verify the quality of bond
between FRP and the substrate, before and after the conduction of the
described semi-destructive tests. For instance, some tests have been
conducted on a reinforced concrete structure on which FRP strips have
been previously applied for shear tearing tests. From the thermograms,
a crack could be noticed corresponding to the strip subjected to the
semi-destructive test. Such a technique is then useful for the detection
deterioration and damage in the structures.
Influence of roughness surface on the debonding force of FRP
In order to investigate the effect of the concrete surface preparation
method on the roughness surface and debonding force of FRP, an experimental campaign has been carried out. The specimens have been produced with different formworks (staves, panels) and different compaction types (beating, vibration). Moreover, two different concrete
strengths have been used in realizing the specimens in order to evaluate also their influence on the surface preparation method efficacy.
Thirty 15 cm-length standard cubes have been also poured and used to
evaluate the mechanical properties of concrete (according to Italian
standards). Mean compressive strength (Rcm = 15 or 20 MPa) from the
compression tests has been obtained by standard cube at an age of 50
days, corresponding to same period of the first profilometer tests. The
number of specimens considered in the present experimental campaign
and the two different casting processes are shown in Table 1.
After curing, four different methods for surface preparation have been
applied on concrete prisms in order to study the effect of the treatment
on the FRP-concrete bond strength: Grinding – the upper surface of the
concrete block has been grinded with a stone wheel to remove the top
layer of mortar, just until the aggregate was visible; Sand Blasting – the
concrete surface has been sand blasted in order to remove the whole
mortar over the aggregates, so obtaining a very rough concrete surface;
Brushing – the surface has been brushed with a twisted steel cord bonded to a rotating disc; Scabbling – impacting the substrate at variable
angle with a metallic tip to create a chipping and powdering action. The
driving mechanism is compressed air.
In order to examine exhaustively the concrete surface, an extensive
campaign of laser profilometer analysis was carried out before and after
RESEARCH - Seismic behavior
the surface preparation. The profilometer used is the DRSC produced
by Miami University, which gives quality and quantity information. This
instrument using laser striping, highlights the rough concrete surface by
thin slits of red laser light at an angle of 45 degrees, and the surface is
observed at 90 degrees by an high-resolution (tiny) board CCD camera
(Figure 38a). The video image of the laser stripes is digitized with a
PCMCIA frame-grabber. The projected slit of light appears as a straight
line if the surface is flat, and as a progressively more undulating line as
the roughness of the surface increases. Lasers with one to eleven stripes
were used.
The roughness degree can be identified by several parameters obtained
by laser profilometry and each one can give specific information on
roughness and its particular properties. They can be classified in amplitude parameters (Ri of Figure 37) and slope parameters (ia of Figure
37a). The first group is sensitive to roughness morphologies, where the
surface is either stepped or slotted and might be described as a discontinuous roughness, the second group is insensitive to roughness morphologies and is more useful for characterizing continuous roughness.
Table 1 Specimens and summary of test sample
N° Specimen
Mean Compressive
Strength (N/mm2)
20 ?
160 x 400 x 600
Type of
(wooden) Compaction
Fig. 37- Roughness parameters.
Figure 37b shows some output parameters provided by the profilometer: Rmax – maximum peak and valley, rough measure of the vertical distance between the highest peak and the lowest valley; Re measures the
vertical distance between the highest peak and the centerline of the
profile; Rv measures the vertical distance between the lowest valleys
and the centerline of the profile; R measures the average of all individually measured peak to valley heights, Rp – roughness profile index,
defined as the ratio of the true length in the fracture surface trace to its
projected length in the fracture plane; iA is the micro-average inclination angle, representing the average of the pixel to pixel angles of the
stripe profile.
In order to define a unique parameter for describing the surface roughness, profilometer parameters were analyzed and correlated. For each
parameter given by the profilometer, average value and covariance have
Fig. 38- (a) Laser profilometer, (b) geometry of FRP-strengthened prism and (c) experimental set-up.
Fig. 39- Average values (a) and covariance (b) of IR parameter for different surface preparations.
been calculated for evaluating the quantity and quality factors of roughness. The covariance can provide for information on surface homogeneity; both quantities are interesting especially regarding the efficiency of
concrete surface preparation methods. In the following, the roughness
is described by coefficient IR = Ria, where R and ia have been
described before. The parameter IR is used to give information on the
absolute value of the roughness and on its specific shape. Profilometer
analysis allows to correlate the casting methods with various degrees of
roughness. In fact, the specimens casted with staves are more rough
because the disconnection of the staves increase the surface irregularity. The specimens compacted by means of the vibration are more rough
due to different positions of aggregates and the presence of vacuum produced by air bubbles.
The roughness of the concrete surface has been investigated before and
after the preparation; its difference is a way to evaluate the efficiency of
each surface preparation method. In Figure 39a,b are shown the average values and the covariance of the IR parameter for all the different
surface preparation methods. Figure 39a shows that all the surfaces
prior to the treatments have very similar roughness while after them the
mean value of the IR parameter is particularly high in the case of scabbing and sand-blasting. On the contrary, all the surface preparation
methods strongly reduce the statistical dispersion of IR parameter
(Figure 39b); the scabbing and sand-blasting provide for the higher
level of homogeneity.
been conducted on two half bricks placed inside a properly designed
steel frame and connected between them by means of two carbon, glass
or natural FRP strips glued on both sides of the specimens. The frame
has been designed in such a way as to avoid any hindrance to the specimen collapse. The specimens have been subjected to uniaxial tensile
tests under displacement control by means of an electro-mechanic testing machine, with a capacity of 100kN, connected to a personal computer. Some specimens have been also monitored by means of unidirectional strain gauges applied both on the brick surface and on the
From the experimental tests, the specimens failure mode has been
analysed. Collapse has occurred for delamination in almost all cases.
However, some specimens have reached failure due to fibres crack, as in
the case of glass fibers and natural ones due to the different stiffness of
the FRPs. More than 30 specimens have been tested. In the case of specimens reinforced by carbon and glass fibers, the load-displacement
curve relative to failure for delamination obtained from the strain gages
applied on FRP and brick surfaces was almost linear until failure which
occurred suddenly and for debonding of the fabric from the reinforced
bricks, with consequent removing of part of the substrate surface. The
presence of relevant traction stresses is observed at the attachment of the
FRP strip with the bricks where the delamination phenomenon begins.
For what concerns the tests conducted on bricks reinforced by natural
fibers, the results, obviously, cannot be compared to the ones derived
from brick reinforced by CFRP or GFRP; however, the tests have shown
interesting properties of natural fibers for not bearing applications such
as on ancient masonry structures where FRP mechanical behaviour
sensibly affects the global behaviour of the structure.
Finally, for each tested specimen, an accurate study of the substrate
after failure has been conducted for the exact definition of the delaminated surface dimensions; the crack begins at a depth of about 10 mm
Delamination tests
Delamination tests have been carried out on bricks reinforced with FRP
materials, with the aim of studying both the collapse load value and the
failure modes that can occur if FRP materials applied during strengthening interventions collapse. In particular, delamination tests have
Fig. 40- Failure mode of bricks reinforced with CFRP strips subjected to delamination tests.
RESEARCH - Seismic behavior
and this distance obviously depends on the experimental equipment,
whereas the depth and the form of delaminated substrate appear almost
constant and dependent on the testing modality (Figure 40).
From both in-situ and laboratory tests, important information were
obtained regarding composite materials behaviour and experimental
procedures. In particular, it was noticed that shear tearing tests are
extremely affected by instrumental errors that can take place during the
test conduction. In fact, while FRP laminates allow a perfect alignment
of the composite with respect to the hydraulic pull machine, in the case
of FRP strips the applied force is not perfectly aligned with the FRP –
concrete interface, and peeling stresses are generated, so causing a significant reduction of the debonding force.
Then, on the basis of the above described experimental results and in
accordance to workshops organized during the Research Activity, some
variations to the CNR DT 200/2004 guidelines, with particular regard
to shear tearing limit value and experimental procedure, were proposed.
In particular, shear tearing test should be carried out following two different procedures, namely direct and indirect procedures. The direct
procedure is preferred if the load can be directly applied on the FRP
glued onto the specimen surface, especially in the case of laminates. If
the FRP system is made of strips to be prepared in-situ, the indirect test
procedure is preferred, in which load is applied by means of steel plates
glued onto the FRP surface.
In this research task a large number of experimental tests has been
performed in order to validate the design equations provided by CNRDT200/2004 and to calibrate some coefficients included in these
equations to better fit theoretical predictions and experimental
Experimental tests have been carried out on both real scale or scaled
members by using traditional FRP systems or innovative typologies of
FRP systems and advanced materials. The experimental tests have
RESEARCH - Seismic behavior
been focused on both masonry and reinforced concrete members. In
particular, advanced FRP materials made of basalt fibers (bonded
with traditional or inorganic cement matrix) have been used for the
confinement of RC and masonry columns. In addition, the mechanical
properties of several FRP systems at fixed environmental conditions
and/or under cyclic actions, the behaviour of RC members confined or
strengthened in flexure and shear as well as of beam-column and
beam- foundations joints have been deeply analyzed. Design criteria
for the seismic retrofit of masonry members and structures have been
also provided. The outcomes of experimental activities have, in most
cases, confirmed the reliability of CNR-DT200/2004 provisions, especially for RC members. Regarding to masonry members, test results
indicate that some coefficients of theoretical expressions provided by
CNR-DT200/2004 have to be refined and some other experimental
results could be necessary to derive theoretical relationships specifically targeted at different masonry substrates and failure modes.
Innovative materials for the vulnerability mitigation of existing structures have been largely analysed in the research activity. Further, new
typologies of FRP systems, also made of several fibers and matrices
types are spreading; on these new typologies further investigations are
strongly necessary in the future. The study of the feasibility of using
composite systems made of inorganic matrix strengthened with natural
fibers fabrics is needful. In fact, the idea of using natural fibers is due
to economic and environmental sustainability suggested by the use of
such materials. These studies could allow new FRP systems to be
inserted in the actual guidelines in order to increase their use in the
civil engineering applications.
The indications provided by CNR-DT 200/2004 for the vulnerability
mitigation of existing structures have been accepted by recent Italian
guidelines, but further experimental tests will need, especially to mitigate the vulnerability of existing masonry structures.
- Aprile A., Benedetti A., Cosentino N., (2006), “Seismic Reliability of
Masonry Structures Strengthened with FRP Materials”, 100th
Anniversary Earthquake Conference, San Francisco, paper n° 1677.
- Aprile A., Benedetti A., Steli E., Mangoni E., (2007), “Seismic Risk
Mitigation of Masonry Structures by Using FRP Reinforcement”,
FRPRCS-8 University of Patras, Patras, Greece, paper n° 424.
- Anselmi V., Aprile A., Benedetti A., (2005), “Safety and reliability of
structures including ductile and brittle elements”, ICOSSAR 2005,
Augusti, Schuëller, Ciampoli (eds), pp. 2183-2188, Rotterdam, ISBN
90 5966 040 4.
- Ascione F., Feo L., Olivito R.S. and Poggi C., “La qualificazione dell’esecuzione degli interventi di rinforzo strutturale con FRP a margine
delle recenti ”Istruzioni per la progettazione, l’esecuzione ed il controllo di interventi di consolidamento statico mediante l’utilizzo di compositi FRP””, Proceedings of National Italian Conference “Ambiente e
Processi Tecnologici – La certificazione di Qualità dei materiali e dei
prodotti da costruzione” (in Italian), Naples (2005).
- Bastianini F., Olivito R.S., Pascale G. and Prota A. (2005), “Controllo
di qualità e monitoraggio dei rinforzi in FRP”, L’Edilizia (in Italian),
139, 66-71.
- Benedetti A. and Steli E. (2007), “Analytical Solution of the Shear –
Displacement Curve for Reinforced Masonry Panels”, The Tenth North
American Masonry Conference, St. Louis, Missouri, ISBN 1-929081-286.
- Benedetti A., Camata G., Mangoni E., and Pugi F., (2007), “Out of
Plane Seismic Resistance of Walls: Collapse Mechanisms and Retrofit
Techniques”, The Tenth North American Masonry Conference, St. Louis,
Missouri, ISBN 1-929081-28-6.
- Benedetti A., Mangoni E., Montesi M, Steli E. (2007), “Verifiche di
Sicurezza ed Interventi di Consolidamento Della Chiesa di S. Martino
in Casola”, INARCOS, 680, pp. 411-423.
- Benedetti A., Steli E., (2008), “Analytical models for shear–displacement curves of unreinforced and FRP reinforced masonry panels”,
Construction and Building Materials, 22, pp. 175-185,
- Bianco V. (2008), “Shear strengthening of RC concrete beams by
means of NSM CFRP strips: experimental evidence and analytical modeling”, PhD Thesis, Dept. of Struct. Engrg. And Geotechnincs, Sapienza
University of Rome, Italy, submitted on December 2008.
- Bianco V., Barros J.A.O., Monti G. (2009a), “Bond Model of NSM FRP
strips in the context of the Shear Strengthening of RC beams”, ASCE
Journal of Structural Engineering, in press.
- Bianco V., Barros J.A.O., Monti G. (2009b), “Three dimensional
mechanical model for simulating the NSM FRP strips shear strength
contribution to RC beams”, Engineering Structures, Vol. 31 n. 4, Elsevier.
- Bruno D., Greco F., and Lonetti P. (2005), “A 3D delamination modelling technique based on plate and interface theories for laminated
structures”, European Journal of Mechanics A/Solids, 24, 127-149.
- CNR DT 200/2004, “Guide for the Design and Construction of
Externally Bonded FRP Systems for Strengthening Existing
Structures”, Italian National Research Council, Rome (2004).
- De Lorenzis L., Rizzo A., (2006), “Behaviour and capacity of RC
beams strengthened in shear with NSM FRP reinforcement”, 2nd Int. fib
Congress, Napoli-Italy, June 5-8, Paper ID 10-9 in CD.
- De Lorenzis L., Galati D. (2006), “Effect of construction details on the
bond performance of NSM FRP bars in concrete”, Proceedings fib
Congress, Napoli, Giugno 2006.
- Dias S.J.E., Bianco V., Barros J.A.O. (2007), “Low strength concrete
T cross section RC beams strengthened in shear by NSM technique”,
Workshop Materiali ed Approcci Innovativi per il Progetto in Zona
Sismica e la Mitigazione della Vulnerabilità delle Strutture, University
of Salerno, Italy, 12-13 February.
- Galati D., De Lorenzis L. (2006), “Experimental study on the local
bond behavior of NSM FRP bars to concrete”, Proceedings CICE 2006,
Miami, USA, December 2006.
- Galati D., De Lorenzis L. (2008), “Effect of construction details on the
bond performance of NSM FRP bars in concrete”, Advances in
Structural Engineering, Multi-science, in stampa.
- Milani G., Rotunno T. , Sacco E. and Tralli A. (2006), “Failure Load
of FRP strengthened masonry walls: experimental results and numerical models”, Structural Durability & Health Monitoring, 2 (1), 29-50.
- Olivito R.S. and Zuccarello F.A. (2006), “Indagine sperimentale per il
controllo dell’applicazione di materiali FRP a strutture murarie mediante prove semi-distruttive e non distruttive”, Proceedings of National
Italian Conference on Materials and Structures Experimentation (in
Italian), Venice.
- Rizzo A., De Lorenzis L. (2006), “Analytical Prediction of Debonding
Failures in RC Beams Strengthened in Shear with NSM FRP
Reinforcement”, Proceedings CICE 2006, Miami, USA, December
- Rizzo A., De Lorenzis L. (2007), “Modelling of debonding failure for
RC beams strengthened in shear with NSM FRP reinforcement”,
Proceedings FRPRCS8, Patras, Luglio 2007.
- Rizzo A., De Lorenzis L. (2009a), “Behavior and capacity of RC
beams strengthened in shear with NSM FRP reinforcement”,
Construction and Building Materials, Elsevier, Vol. 23, No. 4, pp. 15551567.
RESEARCH - Seismic behavior
- Rizzo A., De Lorenzis L. (2009b), “Modeling of debonding failure for
RC beams strengthened in shear with NSM FRP reinforcement”,
Construction and Building Materials, Elsevier, Vol. 23, No. 4, pp. 15681577.
RESEARCH - Seismic behavior
- Savoia M., Ferracuti B. and Mazzotti C. (2003), “Delamination of FRP
plate/sheets used for strengthening of r/c elements”, Proceedings of
Second International Structural Engineering and Construction (ISEC02), Rome, Italy.
Experimental Research on Seismic
Behavior of Precast Structures
ver the last two decades an extensive experimental and theoretical research activity aimed to investigate the seismic behaviour of precast structures has been carried out at European scale.
The results of this activity allowed to consolidate a good knowledge of
the seismic behaviour of precast structures and contributed to the
achievement of prefabrication in Europe with outstanding realizations
in terms of both quality and reliability.
The research was developed within six research programmes. The
first stage developed between 1992 and 1996 during the drafting of
the first ENV version of Eurocode 8 (EC8). The initial draft of the specific rules for precast structures gave them the presumption of a very
Dimensionless values
Fig. 1- Prototype of precast column.
bad seismic behaviour. Industrial one storey precast buildings were
defined as inverted pendulum systems to which a very low value of the
behaviour factor was recognised. This was in conflict with the national codes that did not make any difference between cast-in-situ and
precast frames. This was in conflict also with the experience of past
earthquakes, where precast structures, except for the dry supports,
showed a very good ductile behaviour despite their non seismic design. The lack of experimental data was adduced to justify the heavy
penalisation of precast structures in this code.
In order to verify the validity of this penalisation the Italian
Association of prefabrication industry promoted a campaign of analytical investigations that have been developed at Politecnico di Milano.
The investigation have been started with a set of cyclic tests on precast columns. Figure 1 shows one of the prototypes of precast columns
tested at ELSA Laboratory of the Joint Research Center of the
European Commission at Ispra, Italy. Cyclic and pseudodynamic tests
were performed for different reinforcement amount and axial actions.
An example of force-displacement diagrams obtained from cyclic
tests is shown in Figure 2.a. The energy dissipated over the halfcycles was compared with the maximum value of dissipated energy
associated to a perfect elastic-plastic cycle, as shown in Figure 2.b.
The results of these tests confirmed:
- a good ductile behaviour with specific dissipation around 0.4, as
typical for cast-in-situ columns;
- a more reliable behaviour due to the absence of bar splices and to
the stable position of stirrups during the casting of concrete (precast
columns are cast in an horizontal position);
Fig. 2- (a) Force-displacement diagrams obtained from cyclic tests. (b) Energy dissipated over the half-cycles compared with the maximum value of dissipated energy associated to a perfect elastic-plastic cycle.
RESEARCH - Seismic behavior
Fabio Biondini, Giandomenico Toniolo
Department of Structural Engineering, Politecnico di Milano, Milan, Italy
Fig. 3- Energy dissipation in one-storey frames: (a) monolithic and cast-in-situ; (b) hinged and precast.
- the fundamental importance of a narrow spacing of the stirrups
against the early buckling of longitudinal bars;
- displacement ductility ratios between 3.5 and 4.5 consistent with the
code provisions for cast-in-situ frames.
The ENV EC8 was therefore published with precast structures no
more treated as inverted pendulum, but still penalised with a lower
behaviour factor with respect to cast-in-situ systems.
The seismic behaviour of cast-in-situ and precast structures has been
also investigated by means of proper numerical models on probabilistic bases. This investigation was carried out by means of non linear
dynamic analyses reproducing the real vibratory behaviour of the
structures under earthquake conditions. The aim was to demonstrate
that, under the same seismic action, the monolithic frame shown in
Figure 3.a, with four critical cross-sections dimensioned for a moment
mFh/2, may dissipate the same amount of energy which the hinged
arrangement shown in Figure 3.b dissipates in its two critical crosssections, dimensioned as they are for a double moment M=Fh2m. In
fact it is the global volume involved in dissipation, and not the number of plastic hinges, that gives the total amount of energy dissipated
by the structure.
To demonstrate this assumption, the two prototypes shown in Figure 4
were considered, the first cast-in-situ with monolithic connections,
the second precast with hinged connections. They have the same overall dimensions, with the size and reinforcement of the columns chosen to achieve the same vibration periods and the same design seismic capacity in terms of base shear strength. With the combination
of the different heights and cross-sections (Figure 5) a number of
types of frame were selected for both the frames.
The seismic response of the prototype was investigated in probabilistic terms for lognormally distributed material strengths and under
artificial accelerograms, randomly generated so to comply with the
design response spectrum. A Monte Carlo simulation based on a large
Fig. 4- Prototypes of one-storey frames: (a) monolithic and cast-in-situ; (b) hinged and precast.
Fig. 5- Cross-sectional details of the columns.
RESEARCH - Seismic behavior
Fig. 6- Statistical distribution of overstrength ratio k: (a) monolithic and cast-in-situ; (b) hinged and precast.
Fig- 7- View of the structural prototypes with (a) monolithic and (b) hinged beam-column connections.
sample of incremental nonlinear dynamic analyses taken up to collapse was therefore carried out for each prototype to compute the statistical parameters of the overstrength , ratio of the computed value
over design value of the seismic capacity. Figure 6 shows the distribution of overstrength computed for the two prototypes for a set of
1000 accelerograms in one of the cases studied. These results prove
that precast structures have the same seismic capacity of the corresponding cast-in-situ structures, and confirm the correctness of the
values given by the code to the behaviour factor of concrete cast-insitu frames (q=4.5).
Fig. 8- Displacement time-histories for one of the pseudodynamic tests: numerical (thick lines) versus experimental results (thin lines).
RESEARCH - Seismic behavior
Fig. 9- View of the structural prototypes with roof elements with axis (a) parallel to the direction of the seismic action (Prototype 1), and (b) orthogonal to the direction of the seismic action (Prototype 2).
The third stage of research developed during the revision of EC8 for
its conversion to the final EN version. The preceding analytical
demonstration was effective, but an experimental confirmation was
still necessary. Therefore, taking advantage of the Ecoleader programme for the free use of the large European testing facilities, two
pseudodynamic tests on full scale prototypes have been performed at
ELSA Laboratory. The aim was the direct experimental comparison
of the seismic capacities of cast-in-situ and precast structures, and
at the same time the validation of the analytical model used in the
previous numerical investigation. Figure 7 shows a view of the full
scale prototypes. The comparison of the results obtained for these
prototypes highlights the expected large seismic capacities (about
ag=1g) of this type of structures and confirms the overall equivalence of the seismic behaviour of precast and cast-in-situ structures (see
Fig. 10- Experimental tests carried out on a connection. (a) Test set up. (b) Failure mechanism.
RESEARCH - Seismic behavior
Fig. 11- Tree-storey full-scale prototype. (a) Transversal and (b) longitudinal section.
Figure 8).
The fourth stage of research was developed within the Growth programme. Two further prototypes consisting of six columns and a mesh
of beams and roof elements were designed to investigate the seismic
behaviour of precast structures with roof elements placed side by
side. Figure 9 shows a views of these prototypes and of the testing
plants. The prototypes differ only for the orientation of the beams and
roof elements with respect to the seismic action. Common hinged connections are used between roof elements, beams and columns. The
control of the pseudodynamic test is based on two degree of freedoms,
associated with the top horizontal displacements of the lateral frames,
and of the central frame. Also the effects of cladding panels on the
structural response has been investigated.
The measured top displacements of lateral and central columns
during the pseudodynamic tests resulted practically coincident. This
result proves that double connection between beams and roof elements gives a rotational restraint in the roof plane which enables the
activation of an effective diaphragm action, even if the roof elements
are not connected among them. After the pseudodynamic tests both
prototypes have been subjected to a cyclic test under imposed displacements up to collapse. With a ultimate displacement du360 mm and
a yielding displacement dy80 mm, a global displacement ductility
equal 4.5 is deduced, as assumed by the new final version of EC8 for
the behaviour factor of precast frame systems.
The results of the investigations carried out under the Ecoleader and
Growth research projects showed the good seismic performance of
precast structures under condition that the connections are properly
over-dimensioned. The last aspect to be still clarified is therefore the
actual behaviour of connections under seismic excitation. Based on
these needs, the European research program Safecast has been recently launched to investigate the seismic performance of connections in
precast systems. This project will involve a large campaign of experimental static tests carried out on single specimens, such as the connection shown in Figure 10, as well as pseudo-dynamic tests on a
three-storey full-scale prototype shown in Figure 11. The results of
the Safecast project are expected to complete the large research program developed in Europe over the last two decades which provided
significant advances in the understanding of the seismic behaviour of
precast systems and in the definition of reliable design criteria for this
type of structures.
A number of partners participated to the research, coming from the
principal European countries subjected to seismic hazard. The national associations of prefabrication industry were involved (ANIPB for
Portugal, ANDECE for Spain, ASSOIBETON for Italy, SEVIPS for
Greece, TPCA for Turkey). Also the Italian associations of cement
industry AITEC and ready-mix concrete ATECAP participated to
some stages. The “research providers” were: JRC – Joint Research
Centre of Ispra (of the European Commission), LNEC – Laboratorio
Nacional de Engenharia Civil of Lisbon, Politecnico di Milano,
University of Ljubljiana, NTUA – National Technical University of
RESEARCH - Seismic behavior
Athens, ITU – Istanbul Technical University, and the private laboratories LABOR and LUGEA (I). The “users” were some producers of
precast structures (Magnetti Building – I, Gecofin – I, Civibral – P,
Prelosar– E, Proet – GR), and some auxiliary companies (Halfen – D,
DLC – I).
What described in the present report refers to the series of European
RESEARCH - Seismic behavior
researches coordinated by prof. Giandomenico Toniolo of Politecnico
di Milano. Many other theoretical and/or experimental researches
have been performed on precast structures by initiative of single companies for their specific interests. And this wide activity qualifies the
prefabrication industry as one of the most advanced sectors in growth
and innovation.
State-of-the-art on the Research on
Structural Concrete in Italy
he research activity carried out in Italy on the structural concretes
is very busy in both the academic compartments and in the industrial operations. The present report summarizes some of the most important works in this area including three aspects of this activity:
- shrinkage compensating concrete in the absence of wet curing;
- properties of concretes with recycled aggregates;
- use of bottom ash from municipal solid wastes incinerators.
lary tension caused by the formation of water menisci developed in capillary pores and responsible for the shrinkage of the cement paste
(Figure 1).
Shrinkage-compensating concretes have been extensively used in the
last forty years to minimize cracking caused by drying shrinkage in
reinforced concrete structures.
The first and most diffused system to produce shrinkage-compensating
concretes involves the use of expansive cements, according to ACI 22398, instead of ordinary portland cement. All these special binders are
based on a controlled production of ettringite.
Another effective method to produce shrinkage-compensating concretes, not covered by ACI 223-98 but commonly used in some countries, like Italy or Japan, lies in the use of a CaO and/or MgO based
expansive agent. This technology seems to be more advantageous with
respect to that based on the ettringite formation from an economical as
well as from a practical point of view.
Recently, the addition of a shrinkage-reducing admixture (SRA) has
been found to improve the behavior of CaO based shrinkage-compensating concretes especially in the absence of an adequate wet curing [1].
Although the actual cause of this synergistic effect has not been completely explained, the use of this technology in construction industry
has been increased, in the last five years, particularly in Italy, with very
interesting results.
In the present report three remarkable examples of special reinforced
concrete structures are presented in which the use of CaO-SRA based
shrinkage-compensating concretes was successfully carried out in order
to prevent shrinkage related cracks and/or joints excessive opening in
the presence of adverse curing conditions which are normally not suitable for the use of this technique.
SRAs (Shrinkage-Reducing Admixtures), are generally based on propylene-glycol ether, neo-pentyl glycol or other similar organic substances,
that are able to reduce the drying shrinkage of concrete up to 50% if
used in 1-2% by mass of cement.
According to Berke et al. [2] the effectiveness of SRA must be ascribed
to the decrease in the surface tension of water .This reduces the capil80
• Fig. 1- Water menisci interact with C-S-H fibers determining the shrinkage on cement paste,The New Concrete.
Recently [3], the combined addition of a shrinkage-reducing admixture
with a CaO-based expansive agent has been found to be very successful in producing restrained expansion of laboratory specimens protected from water evaporation for just 1 day by using a plastic sheet and
then exposed to air (60% R.H).
The influence of the SRA on the length change behaviour of a shrinkage-compensating concrete includes two different aspects:
- the effect in Fig. 2 due to a reduction in shrinkage when the concrete is exposed to drying, as expected for the presence of a shrinkagereducing admixture;
- the unexpected effect, which is an increase in the restrained expansion when the concrete is protected from drying with respect to that
obtained without SRA, all the other parameters being the same.
By using a combination of CaO and SRA, then, it is possible to reduce
the amount of expansive agent needed to obtain a fixed restrained
expansion. This reduces the risk of residual un-reacted lime in the concrete.
Furthermore, the performance in terms of initial restrained expansion
and final restrained shrinkage (or residual expansion), of SRA + CaObased shrinkage-compensating concretes is less dependant on the curing efficiency so that the practical use of this technique is easier and
the results are more reliable.
The synergistic effect in Figure 2 has been confirmed by Maltese et al
[4] who have found that the use of a CaO-based expansive agent with a
shrinkage reducing admixture allows to obtain mortars less sensitive to
drying. These authors hypothesize that the synergistic effect of the
RESEARCH - Concrete
Mario Collepardi*
* ENCO, [email protected]
SRA-CaO combination must be ascribed to the massive formation of
CaO elongated crystals during the first hours of curing.
The same authors in [5] propose another mechanism of action: since the
SRA is an organic hydrophobic molecule, it could reduce the water solubility of CaO, retarding its reaction and, then, increasing the
restrained expansion according to Chatterji [6].
Otherwise, Tittarelli et al. [7] have found that SRA doesn’t affect the
speed of CaO reaction with water.
• Fig. 2- Schematic view of the influence of SRA on the length change behavior of a shrinkage-compensating concrete.
Although this synergistic effect has been confirmed by several authors,
the actual mechanism of action needs further investigations in order to
be completely understood.
Notwithstanding this lack of knowledge, the use of this technology, in
the construction industry, has been growing in the last 5 years with
many successful and very interesting results.
In the second part of this paper, three remarkable case histories of special reinforced concrete structures are presented in which the use of
CaO + SRA-based shrinkage-compensating concretes was successfully
carried out in order to prevent shrinkage-related cracks and/or joints
excessive opening in the presence of adverse curing and thermal conditions.
The difficulties encountered in using this technique, in each case, will
then highlight describing the countermeasures which have been taken
to overtake them.
London, U.K.) had proposed the construction of several architectural
concrete walls (20 meters high and 60 meters long) having a sinuous
shape and no contraction joints (Figure 3).
A special CaO-SRA based shrinkage-compensating self-compacting
concrete (SCC) was studied in order to assure a marble-like look, as
required by the designers, even in the presence of a very congested
reinforcement (Figure 4) and, in the same time, to avoid the formation
of shrinkage related cracks along the surface.
• Fig. 3- View of bent and joint-less walls of the MAXXI, Rome, Italy.
2.1 Case History of shrinkage compensating concrete in the absence of
wet curing: MAXXI of Rome
The Museum of Arts of XXI century (MAXXI) in Rome was the first relevant Italian construction in which a SRA + CaO-based shrinkagecompensating concrete has been used (2004-2006).
For this very prestigious building, the designers (Zaha Hadid Limited,
RESEARCH - Concrete
• Fig. 4- Example of steel congestion in a typical wall of MAXXI, Rome, Italy.
Figure 5 shows the strength development with time of the three compared SCCs (CaO-SRA, only CaO and Plain). The strength of the expansive concretes was higher than that of the plain mix. This is probably
due to the consumption of a small part of mixing water caused by the
transformation of CaO into Ca(OH)2 which happens when the concrete
is still in the plastic state and to the consequent reduction of the actual
On the other hand, a slight decrease in the compressive strength of the
SRA + CaO mix was recorded if compared to that of the CaO mix due
to the presence of SRA as experienced in [8].
Although it was specified to protect the concrete surface for at least
three days (to assure a correct hydration of the concrete cover) shrinkage compensating concrete was designed in order to warranty a residual restrained expansion of about 200 mm/m even in case of deficient
curing consisting in just 24 hours of protection by the formwork.
Figure 6 shows the length change of the reinforced prismatic specimens
manufactured with the three different SCCs according to ASTM C 878.
Specimens were not put under water for 7 days as specified in ASTM C
878 test method but were protected with a plastic film for just 24 hours
(to simulate the protection offered by the formwork) an then exposed to
unsaturated air (60% R.H.) at 20°C.
This curing condition was later introduced as “curing method B” in the
last version of the Italian standard UNI 8147 in addition to the “curing
method A” previously specified, consisting in a total immersion in
water for 7 days as in ASTM C878. Actually, the curing method B
appears to be more realistic and similar to jobsite conditions.
Even under these un-favourable conditions of curing, the CaO-SRA
shrinkage-compensating concrete performed very well since the
restrained expansion after 24 hours of protection with a plastic film was
as high as 560 mm/m and, even after 140 days of exposure to unsaturated air, a residual restrained expansion of about 250 mm/m was
recorded. On the contrary, the conventional CaO-based shrinkage compensating concrete showed a lower initial expansion (at lest 320 mm/m)
which completely disappeared after a week of exposure to air after
which, the concrete started to shrink.
Obviously, the plain concrete showed the worst performance reaching a
restrained shrinkage of about 550 mm/m after 60 days when some
cracks appeared on the specimen surface.
Comparing the behaviour of the CaO+SRA-based mix to that of the conventional shrinkage-compensating concrete, both the and effect of
Figure 6 can be detected.
On the basis of the above results, the customer and the contractor
decided to adopt the SRA+CaO-based shrinkage-compensating SCC
• Fig. 5- Strength development of three different SCCs.
• Fig. 6- Length change with time of the three different SCCs.
In order to demonstrate the effectiveness of this type of concrete in offset the formation of shrinkage cracks, its performances were compared
to those of an ordinary CaO-based shrinkage-compensating concrete
(without SRA) and of a plain SCC mixture without expansive component and SRA.
Table 1 shows the composition of these three SCCs having the same w/c
(0.48) and approximately the same cement dosage (350 kg/m3).
Table 1 – Composition of three different SCC
Cement CEM II A/L 42.5R
Limestone filler (kg/m3)
Gravel 4-16 mm (kg/m3)
Sand 0-4 mm
Water (kg/m3)
Acrylic superplasticizer (kg/m3)
CaO-based Expansive Agent
Viscosity modifier
(*) Blended Portland-limestone cement according to EN 197/1
RESEARCH - Concrete
for the manufacturing of all the architectural concrete walls of MAXXI.
Since it was the first time the contractor used an SCC, it was decided to
carry out several field tests, before starting with the manufacturing of
the actual walls, in order to optimize all the casting procedures and test
the suitability of formwork. It was, then, a good chance to test on a real
scale the effectiveness of the expansive technique.
Two field tests were successfully carried out in March and April of 2004
with no cracks formation in two long minor walls of the basement.
A third test carried out in June in order to verify the behaviour of the
expansive concrete in the presence of high temperature failed since
after two weeks, some cracks appeared on the wall surface. The maximum temperature during the casting operation was as high as 35°C and
checking the transport documents of the trucks mixer it was verified
that, because of the congested traffic of Rome, the time elapsed between
the starting of mixing, in the batching plant, and the casting of concrete
into the forms had been in the range of 60-90 minutes, notwithstanding
the batching plant were located near the jobsite.
For this reason the cause of the failure was ascribed to a combined
effect of the high temperature and of a too prolonged mixing time. This
hypothesis was confirmed by laboratory tests in which some ASTM
C878 prismatic specimens were manufactured at 20°C (with raw materials kept at 20°C for 24 hours before the use) whereas other similar
specimens were manufactured at 30°C (with raw materials kept at 30°C
for 24 hours before the use). In both cases, some specimens were put
into the forms after 5 minutes of mixing whereas the others were kept in
the mixer (in movement) for 60 minutes before casting at the same temperature of manufacturing (20 or 30°C). After setting time (about 6
hours) the specimens were demoulded and protected with a plastic film
till 24 hours, at the same temperature of manufacturing (20 or 30°C).
Successively, the specimens were exposed to unsaturated air (60%
R.H.) at the temperature of manufacturing (20 or 30°C).
Figure 7 shows the behavior, in terms of restrained expansion or shrink-
• Fig. 7- Restrained expansion or shrinkage in different manufacturing and curing condition.
RESEARCH - Concrete
age, of the various specimens manufactured.
As expected, the specimens manufactured and cured at 20°C performed well showing a residual restrained expansion in the range of
210÷280 mm/m after 28 days of exposure.
The prolonged mixing (at the same temperature of 20°C) caused a
decrease of the initial as well as in the residual expansion as reported
in [2].
A little higher decrease was recorded in the expansion of the specimens
manufactured and kept at 30°C and cast after 5 minutes of mixing.
Anyway the behaviour of these specimens can be considered acceptable.
On the contrary, the combination of a high temperature of manufacturing and curing and a prolonged mixing cause a strong reduction in the
initial restrained expansion which was completely cancelled after just
one week after which the concrete started to shrink.
The problem was not eliminated by increasing the amount of expansive
agent up to 45 kg/m3 so that, being impossible to assure a transportation time lower than 60 minutes, the contractor decided to delay the
begin of the main wall construction to the autumn and to stop it during
the whole next summer.
The problem of recycling industrial wastes is of vital importance for a
sustainable progress in order to avoid disposals in the environments and
possibly to save resources for the next generations. Such a problem has
already been faced in using fly ash, silica fume and blast furnace slag,
all wastes coming from industries other than cement and concrete.
During the last decade a similar problem has been found for wastes
coming from the construction industry and from the concrete in particular [9] [10].
These wastes can be recycled as aggregates for concretes with two
– first, to save the environment specially in countries, like Nederland
and Belgium, where the available area to build is very limited;
– second, to recycle this waste as aggregate specially in areas where
natural or artificial aggregates are scarce.
Therefore, concrete recycling, by using the readily available concrete as
an aggregate source for new concrete or pavement subbase layers, is
gaining importance because it protects natural resources and eliminates
the need for disposals.
Concrete recycling is a relatively simple process. It involves breaking,
removing, and crushing the existing concrete into a material with a
specified size and quality.
The quality of concrete with recycled aggregate depends on the quality
of the recycled material used, the most important aspect being the origin of the recycled material such as concrete or demolition, the latter
including waste from brick walls and other type of rubbles. The crushed
and sieved material, must be deprived by contaminating products such
as wood, paper, plastic, and bitumen. This recycled material can be
used for pavement subbase layers.
The process of recycling demolished concrete is based on four steps:
– selection of wastes;
– crushing concrete blocks;
– removing of contaminating products;
– mixing with virgin aggregates.
Reinforcing steel and other embedded items, if any, must be removed,
and care must be taken to prevent contamination by other materials, such
as: asphalt, soil and clay balls, chlorides, glass, gypsum board, sealants,
paper, plaster, wood, and roofing materials which can be troublesome.
The mechanical plants where to recycle concrete structures are not very
different from those adopted to treat crushed virgin aggregates.
If the material is devoted to concrete production, further crushing and
sieving are needed before mixing it with virgin aggregate [11].
The crushing characteristics of hardened concrete are similar to those
of virgin rock and are not significantly affected by the quality of the
original concrete. Recycled aggregates can be expected to pass the
same tests required for conventional aggregates. The recycled concrete
can be batched, mixed, transported, placed and compacted in the same
way as conventional concrete. Special care is necessary when using
recycled fine aggregate. Only up to 10% to 20% recycled fine aggregate
is beneficial. The aggregate should be tested at several substitution
rates to determine the optimal rate.
3.1 Properties of recycled fresh concretes
The amount of mixing water of the coarse recycled aggregate is about
5% more with respect to that of virgin aggregate at given size. This
value becomes as high as 15% when the recycled aggregate contains
also the fine fraction. This effect is due to the rough texture of the aggregate and the cement paste surrounding the recycled aggregate.
However, the use of superplasticizers and mineral additions can completely overcome this drawback [12].
3.2 Properties of recycled hardened concretes
Due to the higher porosity, related to the lower density, the recycled
aggregates are responsible for the lower strength of the concrete with
respect to the concrete manufactured with virgin aggregates.
Due to the lower rigidity, recycled aggregates are responsible for the
lower modulus of elasticity of the concrete with respect to the concrete
with virgin aggregates. For the same reason, drying shrinkage and creep
of concretes with coarse recycled aggregates are much higher (25-50%)
with respect to the virgin aggregates. The difference can be still higher
if also fine recycled aggregate is used.
The permeability of a concrete with recycled aggregate is higher than
that of the corresponding concrete at a given water-cement ratio. Again,
the cement paste of the recycled aggregate is responsible for this drawback because the cement paste is more porous and permeable of the
virgin stone.
The frost-resistance of the concrete with recycled aggregate is strongly
reduced by the amount of fine fraction of the recycled aggregate.
Therefore, in concrete exposed to freezing and thawing cycles the fraction of recycled aggregate smaller than 4 mm should be removed.
The fine fraction of recycled concrete smaller than 4 mm is responsible
for all the above mentioned limits of the concrete with respect to the
corresponding concrete manufactured with virgin aggregates. However,
if the fine fraction is ground very finely (smaller than 0.1mm) it can be
used advantageously in manufacturing self-compacting concrete as
filler to improve the cohesiveness of the concrete [13, 14].
Mineral solids in form of fly ash and bottom ash are produced by burning municipal solid wastes in incinerators (MSWI). Fly ash is negligible and it is so chloride-rich that it cannot be used as mineral addition
in cement-based mixtures for reinforced concrete structures.
On the other hand, bottom ash is about 25% with respect to MSWI and
its chloride content is negligible, so that it could be potentially used as
mineral addition for manufacturing concrete mixtures. However, ground
bottom ash (GBA) from MSWI does not perform as well as other mineral additions (silica fume or fly ash produced by coal burning) due to the
presence of aluminium metal particles which react with the lime formed
by the hydration of Portland cement and produce significant volume of
hydrogen in form of gas bubbles which strongly increase the porosity of
concrete and reduce its strength.
Due to this drawback, a new process was developed to completely separate the aluminium metal particles through a mechanical removal of
metals and a special wet grinding of bottom ashes. At the end of the
process GBA was used as an aqueous slurry to replace Portland
Some researches have actually shown the pozzolanic activity of ground
MSWI bottom ashes showing their reactivity with lime or portland
cement clinker [15,16]. Nevertheless, no successful use of MSWI botRESEARCH - Concrete
tom ashes as mineral addition in concrete has been reported, because
of the side effects of this addition. According to Bertolini et al [17], the
main side effect is related to the evolution of hydrogen gas after mixing
due to the presence of metallic aluminium. In the alkaline environment
produced by the hydration of portland cement (pH around 13), corrosion of some metals (mainly aluminium) produces a great amount of
gaseous hydrogen. After placing and compaction of concrete, this gas is
entrapped in the fresh material, producing a network of bubbles that
leads to significant reduction in the strength and increase in the permeability of the hardened concrete.
The present report summarizes the results of a research [18] aimed at
developing suitable treatments to allow the use of MSWI bottom ashes
as mineral additions for the production of structural concrete without
the evolution of hydrogen gas due to the presence of metallic aluminium particles.
Ground bottom ashes (GBA) from municipal solid waste incinerators
(MSWI) were manufactured according to a new technology based on a
high degree of separation of metals including the heavy ones, the wet
grinding process, and other specific technical solutions to completely
remove the aluminium metallic particles. At the end of the process, a
fluid slurry was obtained with particle size in the range of 1-5 mm. By
changing the wet grinding time three GBA were produced with a mean
particle size of 5 mm, 3 mm and 1.7 mm.
Compressive strength and durability measurements were carried out in
concretes where Portland cement was replaced by 20% of ground bot-
RESEARCH - Concrete
tom ashes from MSWI in comparison with concretes containing 20% of
coal fly ash or 10% of silica fume.
The performances of GBA with mean sizes of 3 and 5 mm were higher
than that of the coal fly ash particularly at 1-60 days. The finest ground
bottom ash (with a mean size of 1.7 mm) performs as well as silica fume
in terms of compressive strength, water permeability, chloride diffusion
and CO2 penetration.
These results appear in particular to be very interesting from a practical point of view since it will be possible to manufacture big amounts of
a pozzolanic material as effective as silica fume (which is not available
and very expensive) in producing high-performance concrete in agreement with a sustainable progress for the re-use of waste materials
instead of a their disappointing disposal.
The research on the progress of structural concrete in Italy is very
active. Three sections of this activity have been illustrated in the present report:
– shrinkage-compensating concrete for crack-free structures even in
the absence of a wet curing;
– recycled concrete with special application of a fine powder material
devoted to self-compacting concrete;
– bottom ash from municipal waste incinerators as pozzolanic materials
for high-performance concrete.
MET/ACI International Conference on Superplasticizer and Other
Chemical Admixtures, Sorrento, 2006, pp. 103-115.
M. Collepardi, The New Concrete, pp. 347-362, Second Edition,
Environmental Conscious Materials and Systems for Sustainable
Ed. Tintoretto, Villorba, Italy, 2010.
Development, 2004, pp. 179-186.
N.S. Berke, L. Li, M.C. Hicks and Bal, “Improving concrete per-
MET/ACI International Conference on Superplasticizer and Other
overview”, Materials and Structures, 39, pp. 655-664, 2006.
M. Collepardi, A. Borsoi, S. Collepardi, J.J. Ogoumah Olagot, R.
gress, Naples, ID 6-19, 2006.
V. Corinaldesi, G. Moriconi, “Role of Chemical and Mineral
Compensating Concrete Under Non-Wet Curing Conditions”,
Admixtures on Performance and Economics of Recycled-
Cement and Concrete Composities, 6, 2005, pp. 704-708.
Aggregate Concrete”, Proceedings of the Seventh CANMET/ACI
C. Maltese, C. Pistolesi, A. Lolli, A. Bravo, T. Cerulli and D.
International Conference On Fly Ash, Silica Fume, Slag and
Salvioni, “Combined Effect of Expansive and Shrinkage
Natural Pozzolans in Concrete, Editor V.M. Malhotra, ACI SP
Reducing Admixtures to Obtain Stable and Durable Mortars”,
199, Vol. 2, pp. 869-884, 2004.
V. Corinaldesi, G. Orlandi, G. Moriconi, “Self-compacting con-
C. Maltese, A. Lolli, C. Pistolesi, A. Bravo, and T. Cerulli,
crete incorporating recycled aggregate”, Innovations and De-
“Combined effect of expansive and shrinkage reducing admixture
velopments in Concrete materials and Construction, Proceeding
on microstructure of mortars and concretes”, Proceeding of the
of the International Conference, Editor Ravindra Dhir, Dundee,
International Conference on Durability of Concrete, Editor: V.M.
UK, pp.9-11, 2002.
V. Corinaldesi, F. Tittarelli, L. Coppola G. Moriconi, “Feasibility
S. Chatterji, “Mechanism of expansion of concrete due to the pres-
and Performance of Recycled-Aggregate Concrete Containing
ence of dead burnt CaO and MgO”, Cement and Concrete
Fly Ash for Sustainable Building Development”, Proc. of the
Research 25, 1995, pp. 51-56.
Three-Day CANMET/ACI Int. Symposium On Sustainable
F. Tittarelli, S. Monosi, G. Moriconi, R. Troli, “Effects of a shrink-
Development of Cement and Concrete, SP 202-11, pp. 161-180,
age compensating Admixture in Shrinkage-Compensating
Concretes”, presented in “9th ACI International Conference on
M. Bassan, M. Menegotto, G. Moriconi, “Precast Structural Concrete with Recycled Aggregates”, Proceedings, the 2nd fib Con-
Malhotra, Montreal, 2006, pp. 781-796).
Malhotra, 2003, pp. 37-50.
Cement and Concrete Research 12, 2005, pp. 2244-2251.
J.Zh. Xiao, J.B.Li, Ch. Zhang, “On relationship between the
mechanical properties of recycled aggregate concrete: An
Troli, “Effects of Shrinkage-Reducing Admixture in Shrinkage
formance with Shrinkage-Reducing Admixtures”, 7th CAN-
Chemical Admixtures in Concrete, Berlin, Germany, Ed. V.M.
N. Kashino and Y. Ohama, RILEM International Symposium
K.A. Paine, R. K. Dhir, V.P.A. Doran, “Unprocessed and
Superplasticizer and Other Chemical Admixtures”, Seville, 2009.
processed incinerator bottom ash as a cement bound material” in
M. Collepardi, R. Troli, M. Bressan, F. Liberatore, G. Sforza,
R.K.Dhir, T.D.Dyer, K.A. Paine (Eds.) “Use of incinerator ash”,
“Crack-Free Concrete for Outside Industrial Floors in the Absence
Proceedings of the International Symposium organized by the
of Wet Curing and Contraction Joints”, Suppl. of 8th CAN-
Concrete Technology Unit, University of Dundee (UK) 20RESEARCH - Concrete
lepardi, “MSWI ashes as mineral additions in concrete”, Cement
A. Macias, E. Fernandez, S. Goñi, A. Guerrero, “Valorizacion de
and Concrete Research, 34, 1899-1906, 2004.
las cenizas de inceneracion de residuos solidos urbanos en los
M. Collepardi, S. Collepardi, D. Ongaro, A. Quadrio Curzio, M.
materiales de construccion”, Papel de los sectores cementero y
Sammartino, “Concrete with bottom ash from municipal solid
de la construcciòn en la gestiòn y reciclado de residuos, CSIC,
wastes incinerators”, paper accepted for publication on “Second
Madrid, 2001.
International Conference on Sustainable Construction Materials
L. Bertolini, M. Carsana, D. Cassago, A. Quadrio Curzio, M. Col-
and Technologies (SCMT)”, Ancona, Italy, June 2010.
RESEARCH - Concrete
Damages of L’Aquila Earthquake
his chapter summarizes the main findings of a chapter of the special issue of Progettazione Sismica devoted to L’Aquila earthquake
the author edited and which is about to appear. In particular, in the following the papers of Carocci and Lagomarsino (2010) regarding
masonry buildings, Cosenza et al. (2010) for reinforced concrete buildings, Di Ludovico et al. (2010) about school buildings, Casarotti et al.
(2010) for hospitals, Menegotto (2010), Faggiano et al. (2010) for industrial structures, Dolce et al. (2010) for lifelines, are reported.
Are masonry buildings able to withstand a strong earthquake such as
the one that struck L’Aquila on the 6th of April, 2009? Is it possible to
repair damaged buildings, guaranteeing an adequate safety level to
their inhabitants in an area with such a high level of seismic risk?
These are the questions asked by researchers and government technicians, but especially by those people who lived through this tragic event
and long to see the restoration of historic centers and return to their
homes but are also afraid.
The answer to these questions must be complex and detailed because
there were so many factors that influenced the seismic behavior of
Aquilan constructions.
In this earthquake, more than others, the effects of localized seismic
amplification played an important role. If one analyzes the macro-seismic consequences of many of the historic centers in the Aterno Valley
(south-east of Aquila), one will immediately observe that the villages
that were struck the hardest (Onna and Villa Sant’Angelo, I=9-10) were
near other villages where the damage was limited: Onna (I=9-10) which
is only 1500m as the crow flies from Monticchio (I=6). Near the town of
Villa Sant’Angelo, the distance between the chief town (I=9) and the village of Tussillo (I=8) is less than 700 m. Even within the town of
L’Aquila, there were zones where the damage was clearly concentrated.
Especially in L’Aquila, it was noted that most victims were found in
reinforced concrete (r.c.) buildings (135 versus 52 in masonry buildings). In the historic center, damage to churches was severe, in some
cases with extended collapse, but the increased vulnerability of these
structures is well-documented by history (in the past, churches were
often found to have the highest concentration of victims). Some severe
damage was rather diffuse in apartment buildings, but did not lead to
collapse except for localized cases in loggias or stairwells; almost
always found in abandoned or poorly maintained buildings.
Modern criteria for structural safety, based on tests of differing limit states (performance-based design), state that during a rare earthquake the
construction may undergo severe damage but must avoid collapse in
order to protect the lives of its occupants. The earthquake on April 6 of
last year was of an intensity comparable to that found in the new
Technical Regulations of Construction, and many buildings in L’Aquila
historic center sustained damage compatible to those norms (the accelerograms recorded by instruments present in the area of the “seismic
crater” revealed a spectrum only slightly higher than that in the building plans).
Therefore, one can attest that well-constructed masonry buildings, in
other words those built with connections that allow the structure to act
as a unified organism (clamping between orthogonal walls, junctures
between walls and floors) are able to sustain damage without manifesting fragility collapse.
This article will not deal with monumental structures, but with minor
masonry ones, common in the many historic centers in and around
L’Aquila, where severe damage and collapse was diffuse. Wishing to
take into consideration the number of victims as a parameter of vulnerability, outside of L’Aquila there were 97 victims in masonry buildings,
compared to 14 victims in r.c. buildings (it should be said that in these
towns the percentage of r.c. buildings is much lower than in L’Aquila
and the buildings were also lower).
What are the factors that justify the different behavior between the
apartment buildings in L’Aquila and the buildings in the surrounding
historic centers? Effectively, there are two reasons: construction quality
and the presence of incongruous subsequent restoration.
In the masonry buildings of the historic center of L’Aquila, most of
which were rebuilt after the tragic earthquake in 1703, a series of structural modifications characteristic of the L’Aquila rules of thumb were
easily recognizable: wooden beams (elements built into the wall thickness and connected externally by way of small tie rods), to improve connections between walls; the connection of roofs to the tops of the walls,
utilizing external wooden posts. These rules were adopted even in the
smaller surrounding historic villages, but often with lesser construction
know-how and lower quality building materials.
Even after the earthquake in 1703, the historic centers of the Aterno
Valley were struck by significant earthquakes, in particular, the one in
Avezzano in 1915. In fact, the repairs and seismic reinforcements are
easy to discern (scarp walls, buttresses, and tie rods adjacent to
masonry walls) and in many cases, the partial reconstruction of collapsed portions. These interventions often functioned well, but in others
the vulnerability remained, proving once again how difficult it is to
perform truly efficient seismic restorations and improvements.
As far as recent restorations were concerned, (only in certain cases done
for consolidation purposes) it should be said that while L’Aquila
appears to be better preserved (due to both the large number of protecRESEARCH - L’Aquila Earthquake
Gaetano Manfredi
Department of Structural Engineering, Università degli Studi “Federico II”, Naples, Italy
ted buildings as well as perhaps the number of uninhabited buildings
in certain areas), the smaller historic centers were subject to frequent
changes: enlargements, transformations, added floors, restored floors
and roofs. These interventions were often done with building materials
and techniques incompatible with the original structures: alterations in
weight displacement, differing rigidity between elements, and dangerous increases in mass.
The final important question is if it is possible to repair severely damaged constructions and rebuild collapsed parts of aggregates with traditional construction techniques (i.e. stone), eventually modified on the
basis of the experience of this earthquake. To this end, an examination
of construction damage to masonry is proposed based not only on the
identification of vulnerable areas, but also on resistance (anti-seismic
regulations) as a means of truly learning something from the test of an
2.1 Typological aspects of masonry buildings in the historic centers of
The historic centers of the L’Aquila territory are principally made up of
simply organized masonry buildings. Even the grouping within connected buildings seems to be to simply follow the rules imposed by the orography of the terrain.
Due to the areas extended (including seismic) history, construction
peculiarities and specific characteristics for the arrangement of aggregates for each of these centers have been adopted, and deserve to be
taken into consideration, but will not be dealt with in this article.
Nevertheless, with reference to aggregates, one can affirm that the
urban centers are characterized by smaller dimensioned blocks of buildings that develop along the morphology of the terrain where they were
In the centers located on mountainsides (for example: Fossa, Casentino,
Tussillo, Castelnuovo) or those built on plateaus (for example: Poggio
Picenze, Sant’Eusanio Forconese, Villa Sant’Angelo), both cases are
characterized by sloping terrain. One can observe the two types of typical aggregates: parallel blocks and blocks built orthogonally to the
incline as well as a wide range of variants in between.
Buildings were arranged according to the conformation of the terrain
and the existing buildings, and the relation to seismic vulnerability can
be listed schematically: height differences between walls on opposite
sides of the street; scaling of contiguous building fronts; number and
height of external exposed walls, and the placement of the building in
the aggregate.
In the oldest centers where the building aggregates developed along
side streets, the aggregates are characterized by the continuity of the
RESEARCH - L’Aquila Earthquake
street front and the contextual absence of building compactness in the
depth of the fabricated bodies (for example: San Demetrio Colle).
Such generalized characteristics are largely identifiable in centers
analyzed herein, even if in our conscience it is not possible to establish
a correlation between the type of building aggregate and the damage
sustained. In any case, the observations relative to the scale of the
aggregate are important not only for identifying where the main vulnerability lies and differentiating it from that of single buildings, but especially for establishing the project phases for restoration and reconstruction.
The diffusion of arches placed between facing walls along the streets is
a repeated tendency in these villages subject to earthquakes. However,
it is not possible to prove that the volume of overpasses (built onto the
inhabitable surface area) though widely present in many of these centers (for example: Santo Stefano di Sessanio), is the result of the evolution of the arches, even if it is obvious that both tend to create a rather
good connection between the facing blocks and constrict wall movement where they are located. Street names such as Contrada dell’Arco,
Chiassetto dell’Arco, Via sotto gli Archi, present in nearly all of the centers we visited, indicate that this construction technique dates way
Buttresses, built-on scarp walls and tie rods should be counted among
the pre-modern anti-seismic defenses that systematically characterize
the building range of these centers. Each technique can be identified
by materials analysis, production, function and use after the many
earthquakes that have struck this area throughout history. In fact, many
of these elements have been utilized in restoration phases while in other
cases, they seem to have become the rule of thumb for the partial or total
reconstruction of buildings.
Evidence of the damage done by historic earthquakes is still often visible, such as the often seen out of axis external walls of buildings, and
the presence of stone elements, often hewn after being taken from older
collapsed buildings and reutilized in post quake reconstruction..
Given the close relationship that links these centers to the territory
(agricultural/pastoral economy), the inhabited buildings still show traces of their rural past with ground floors usually for storage or stalls and
the living spaces in the upper floors accessed by means of external
staircases along a balcony.
The structure of these stone stairs is usually surrounded by pillared
masonry that tends to create volume towards the street front; in cases
where the building has another floor, the additional volume is used as a
loggia while access between the two inhabited upper levels is solved by
way of an internal wooden stair.
In some of the centers visited, the houses showed more “urban” characteristics in how they were set up on the street front while at times
maintaining the external stone stairway to access the inhabited upper
floor. In any case, a common peculiarity seemed to be the add-on of
advancing volume on the front of the house for more living space.
Along with the mono-celled buildings on the street front, which we have
referred to up till this point, larger and more complex structures exist
without being large enough to be considered real apartment buildings
(with the exception of sporadic cases in the smallest villages). These
were derived mostly from the fusion of pre-existing smaller buildings
with the addition of an upper floor.
The “stall-hayloft” is another very common type of building which completes the overview of the minor architecture in these centers. Its function determines its appearance, and though it is completely distinct
from the living area, it is often placed contiguously to the house and therefore inserted into the building mesh. The most important difference
from a construction and structural point of view is that these constructions represent the largest part of the wall structure (where some walls
reach up to 10 m in length) with the presence of dilated wall light which
corresponds to the systematic use of roof coverings connected to external walls.
From our observations, it was possible to conclude that up until now, the
wall thickness of the cell walls were adequate for the dimensions of the
buildings (3 levels above ground and usual light openings); in fact these
prove to be between 50 and 70 cm except at ground level or in rare
cases where stone vaults are present where up to 1 m of thickness can
be found. Congruous tapering was also observed in upper floors; scant
wall thickness was observed only on rare occasions, always in the presence of more recent interventions.
As far as horizontal diaphragms are concerned on the first floor (the
division of lower floors used for storage or stalls and inhabited upper
floors), there were always barrel vaults with the generatrix perpendicular to the façade. The construction technique calls for a mesh of rough
hewn stone elements and the presence of compact buttresses.
The upper floors usually have simple wooden flooring. However, brick
covered metal beams were frequently noted, certainly from substitutions in the last century (for example after the Avezzano earthquake).
Metallic beams, hollow flat blocks covered by flooring were more
recent. In some cases, the beams were connected to bars or welded plates and anchored to the masonry.
Another type of horizontal diaphragm is made of thinly layered brick
vaults (two thin layers of interwoven bricks) which have proven to be
very vulnerable; these vaults are also relatively recent (from the end of
the 19th century to the first decades of the 20th century).
The roof coverings are made of wood with the constant presence of
overhangs made of dripstone or wooden elements. In buildings in aggregates, the ridge of the roof is parallel to the street front and may have
one pitch towards the street or two slopes with a central ridge (depending on the configuration of the whole aggregate). The wooden beams
are always positioned without overhang; even in the case of buildings
placed on corners or at the head of a block where the configuration of
the pitch is a triangular pavilion shape (therefore with the presence of
two directions of slope). The attention paid to not creating overhanging
structures and nevertheless the wooden elements were almost always
connected to the walls. The attention to the way rooms were constructed for reducing seismic vulnerability is also observed in many small
details, such as the “light” eaves made from wooden and bamboo structures and the balconies entirely made of wood. Such construction measures were evidently realized in order to limit the lethal effects of damage on the lives of the inhabitants.
To conclude this chapter which focuses on the characteristics of the
buildings in minor centers of the L’Aquila area, it is necessary to mention two aspects that should be considered in the observation and evaluation of damage: the techniques adopted in recent restorations and
the conservation state of the buildings.
In relation to the first of the two aspects mentioned above, it should be
observed that as long as masonry walls were the dominant technique
used for building, transformations on existing buildings took into consideration this vulnerability and often brought about seismic improvements within the limits of the techniques available, eliminating weaknesses and introducing protective measures.
Unfortunately, recent transformations appear mostly linked to damage
after the earthquake. The most common intervention is the substitution
of the original roof covering with a new one which generally followed the
same configuration of the preceding one, but was at times made from an
r.c. structure with hollow or infill panels or with metal beams instead of
wooden ones. The result of these changes with regards to seismic vulnerability was quite often negative.
As far as the second aspect is concerned, it seems clear that the conservation state of the building played an important role in whether or
not it was damaged, but also with reference to the buildings near it in
the aggregate. In fact, in cases where the adjacent buildings were poorly
constructed due to many decades of abandon, there was less damage
than expected due to the stabilizing contribution of the contiguous cells.
In some of the centers visited, the presence of many restoration work
zones was observed (for example at Villa Sant’Angelo).
As far as we could discern, in most cases the techniques these workers
were utilizing were far from those that should have been adopted for the
knowledgeable recovery of historic heritage. On a positive note, also in
view of the reflections in the next chapter, we also encountered that the
restoration and reuse of abandoned masonry structures had been going
on for some time, and was probably linked to a search for available
RESEARCH - L’Aquila Earthquake
living space close to the city center but also due to a slightly higher tourism interest (also foreign).
2.2 Damage observations: vulnerability and resistance
The description of the damage observed will be discussed by topic
according to the most important building aspects (at times re-examining
the topics mentioned above) and by referring to the principal damage
and collapse mechanisms.
Subdivision is necessary in order to shed light on the probable causes
that facilitated or more aptly limited damage. In this way, characteristics of vulnerability will be highlighted as well as the strengths which
should lead to further reflection for future reconstruction.
2.2.1 Structural organization of buildings and position in the aggregate
As far as the placement of the building in the aggregate is concerned, it
has been revealed that the configuration of the corner or the head of the
block has proven to be the most disadvantageous as widely noted before. In Figure 1, we can observe three buildings each placed at the head
of a block. All of them sustained the collapse of the front wall and a
great portion of the two side walls. Good organization and regularity of
the vertical load elements of the building represents a point of strength
with regards to seismic action. The orientation of orthogonal walls to the
façade, the placement and quantity of openings and the connection with
the horizontal elements determines the greater or lesser flexional trim
(vertical and horizontal) of the walls. Irregularity or changes introduced
in the configuration of the whole often proves fatal during an earthquake.
One often noted case is that of the buildings created by the fusion of
pre-existing contiguous buildings and the later addition of one or more
Fig. 1- Tempera: collapse of buildings placed at the head of a linear aggregate.
RESEARCH - L’Aquila Earthquake
floors. At times, the new building eliminated certain walls orthogonal to
the façade yet present in the lower floors when raising the height, thus
altering the passage through transverse walls (and generally rotating the
orientation of the floors).
Figure 2 shows the case of a building where one of the walls orthogonal
to the façade was absent on the top floor, yet traces can be observed on
lower floors. The excessive distance between the transversal walls rendered overturning nearly inevitable in the case of seism.
Such construction defects, observed on many occasions, can have easily
been revealed in the buildings in the historic centers and the vulnerability derived from such configurations found and resolved preventatively.
Fig. 2- L’Aquila: the absence of transverse walls on the upper floor of the building.
The vulnerability derived from the absence of clamping between walls
of buildings built at different times was confirmed (growth phases and
evolution of the buildings in an aggregate). Figure 3 shows the collapse
of the portion of a wall due to construction discontinuity by the simple
placement of a wall next to the pre-existing corner.
Fig. 3- Poggio Picenze: local collapse due to discontinuity.
Beyond typical growth or urban congestion, many violations of building
codes were found in the transformations of aggregates in historic centers. In other words, demolition and reconstruction of weight-bearing
walls in different positions in order to obtain small extensions or reconfigurations of the block produced negative alterations of cells. In these
cases, the original clamping was lost and nearly impossible to restore
on the newly placed walls, and thereby eliminated connecting elements.
In the case shown in Figure 4, a cut off wooden tie rod is visible right
inside the internal wall, probably during restoration of the building or
its apartments. This alteration, which eliminated an anti-seismic construction norm put in place by previous craftsmen certainly contributed
to the collapse that was verified in that building unit.
the house, and consequently, the use of spaced trusses are necessary
(Figure 5-6). From a construction point of view, it is clear that the connections between roof and walls were put into place at the time of original construction as demonstrated by the position of lintels which
necessarily implies their placement before mounting the boards and the
subsequent roof covering. Finally, it should be noted that the use of connecting trusses to the masonry walls by means of wooden lintels was
also consistently found in churches in the L’Aquila region, and therefore one can assume that this technique was generally associated with
construction configurations with great light.
Connections by means of wooden lintels were also observed in larger
and more important buildings like that in Figure 7 which gives an
example of the positive contribution offered by the roof covering. The
corner building shows systematic tie rods between floors and the top of
Fig. 4- Villa Sant’Angelo: violations: the external wooden connection to the metallic tie rod was cut during fusion when the
wall direction was altered.
2.2.2 Roof Coverings
As already mentioned, wooden roof structures generally do not
overhang, demonstrate well-placed orientation, and are usually wellconnected to the masonry walls evidenced by metallic or wooden connections visible from the external walls. Such connections are associated to the very common presence of pavilion type roof coverings or the
triangular portion of the pavilion roof, where the presence of ridge rafters or slope necessitates the organization of the mesh of the wood to eliminate thrust. In roofs whose main direction is limited by masonry
walls they are posed upon, it was not rare to find that an effective connection to the wall rims contributed to limiting the damage and protrusion of the external walls.
The habit of connecting, often by way of wooden lintels, roofs to
masonry walls seems to be linked to the dilated dimensions of the cells;
in fact this modality is constantly found in uninhabited spaces such as
stalls-haylofts. In the latter, the floor area is usually larger than that of
Fig. 5- Villa Sant’Angelo: a wooden truss leans on a wooden lintel.
Fig. 6- Casentino: masonry cyma of a hayloft and the truss touching the wooden lintel (note how the internal connection is
thanks to a board nailed to the actual lintel).
RESEARCH - L’Aquila Earthquake
the wall put into place when the masonry walls were elevated. One can
also observe the presence of wooden tie rods which are connected to the
trusses which create the pavilion roof structure (the presence of a tie rod
may indicate the absence of an orthogonal wall to the façade). In any
case, it is interesting to note how the building has cracked in the floor
with walls (second mode mechanism) while initial cracking is entirely
absent on external walls.
Fig. 9. Villa Sant’Angelo: diffused collapse of the masonry cyma without involving the window openings.
Fig. 7- Poggio Picenze: a small corner building of a block built on a slope.
Below, the Figures demonstrate how damage can be attributed to a lack
or modification of the roof structure. Damage caused by the poor quality or advanced state of decay of the wooden elements of the roof was
very common. This was due to the lack of connection to the masonry
(along with the scarce conditions of placement), the lack of connection
between main and secondary elements of the roof covering, or the bad
conditions of the wooden boards (which if in good condition guarantee
a very limited bracing effect).
In these cases, local collapse was recorded in the tympanum of the cell
walls (Figure 8), in other words extended collapse to the fascia above
Fig. 8- Poggio Picenze: local collapse of tympanum walls due to holes punched in the rafters.
RESEARCH - L’Aquila Earthquake
the external walls without involving the architrave structure of the openings on the highest floor (Figure 9). In both cases, damage may have
been caused by movement relative to the placement of wooden elements
and the effect of the addition of wall openings which weakened wall
Localized damage in the tympanum walls (which is one of the most
commonly found types of limited damage observed in the L’Aquila centers) may be more or less extended depending on the surrounding conditions, where the tympanum wall was placed in the aggregate (differing
height to the contiguous buildings), or in relation to the quality of the
masonry pattern at the top of the wall.
Extensive damage involving the top of the external walls may be attributed to where the wooden mesh elements (primary and secondary)
were not interconnected which led to greater displacement at the top
part of the wall during the quake.
Localized collapse to a portion of the masonry cyma illustrated in
Figure 10 is presumably due to thrust in the rafters of the slope of the
pavilion roof. Localized cracks are also visible near the corners, perhaps caused by horizontal displacement transferred to the walls by the
angular rafters of the pitch. One can also observe the collapse of the
obstruction of an arched opening.
It was dramatic how many cases of damage were unmistakably caused
by the substitution of traditional wooden rafters with structures which
although appeared to imitate the original had vast differences in weight
and rigidity.
Both in the cases of heavy stringcourses as in those where the weight
is due to infill roof panels, the effect produced by the seism was fatal.
The external walls below were seriously damaged and the roof cove93
ring structure remained on top of what remained of the building
(Figure 11-12).
Fig. 10- Vallecupa: localized collapse of a portion of the masonry cyma.
2.2.3 Horizontal diaphragms and vaults
In the case of floors, it was generally observed that modifications played
a negative role in seismic behavior; recent substitution interventions of
original horizontal structures were quite diffused, in particular in the
largest centers, and they seem to have led to the type of damage discussed below.
The use of floors with metallic beams and brick (small brick arches or
thin hollow flat blocks) was common for nearly all of the last century
and may improve the overall behavior in some cases when well-made.
However, such interventions require localized clamping to the masonry
(of less impact than that necessary for the insertion of r.c. stringcourses)
and in a certain increase in the vertical load on the masonry. These
loads inevitably influence one of the masonry parameters favoring
damage to the transverse connections in the wall and phenomena of
localized instability.
In Figure 13, observe how the portion of the wall collapsed from the
ground floor up to the height of the lintel of the window on the top floor.
The collapsed portion was limited to the vertical alignment of two
columns of puncturing and the nesting seems to have been favored by
the presence of an r.c. floor positioned over the original thin brick vault
and by the presence of the addition of a staircase leading to the attic.
Fig. 11- Tempera: the heavy substituted roof caused the crumbling of the walls of the floor beneath it.
Fig. 13- L’Aquila: a second floor and a staircase in r.c. were placed on top of the original one.
Fig. 12- Villa Sant’Angelo: the r.c. roof with infill panels induced the collapse of the masonry walls beneath it.
An analogous effect seems to have resulted from the presence of an r.c.
floor posed in order to consolidate and increase the rigidity of the original floor in metallic beams with thin layered brick vaults (Figure 14).
In this case, the wall collapsed from the base to the summit with the
exclusion of the r.c. stringcourse and the roof structure which remained
in position.
The head of the building had a rounded corner and the collapse of the
RESEARCH - L’Aquila Earthquake
mon horizontal structure for most of the buildings. It is interesting to
observe that in the range of current seismic damage, the vaults reacted
well. In fact, even in the cases involved in extensive collapse, the vaults
of the ground floors appeared intact (Figure 16). This is certainly due to
the correct dimensions of the masonry piers as well as the good placement of the vault within the complete configuration of the building and
the whole aggregate.
Fig. 14- Paganica: a thick cement layer for rigidity was posed on the horizontal diaphragm.
walls may have been determined by the rigidity of the floor panels
which formed a diagonal point which tended to expel at the corner.
The case documented in Figure 15, merits some attention due to the
fact of its common reccurrence within the complete range of damage.
The deformation localized in correspondence to the size of the horizontal diaphragms, absent at the top of the external wall, may indicate
excessive floor weight and/or the lack of any connection between the
horizontal diaphragms such as a barrel vault or flooring built on beams
parallel to the façade.
Fig. 16- L’Aquila: the vaults remained intact notwithstanding the collapse of the facade and the upper horizontal diaphragm.
Both in the role of horizontal diaphragms as well as in false ceilings for
inhabited attic spaces, the brick vaults were not always spared from
damage (Figure 17). This fact is easily proven both in smaller constructions as well as in important and monumental buildings (the damage is visible to these types of structures in the upper floors of many buildings in the historic center of L’Aquila). To the contrary of stone vaults,
the reduced thickness of the brick vaults rendered them more sensitive
to even the smallest movements of the imposing walls. Obviously, for an
accurate evaluation of vulnerability of this type, the many cases where
Fig. 15- Poggio Picenze: evident deformation of the masonry walls of the facade with the worst damage at the level of the
horizontal diaphragm.
As previously mentioned, the use of vaulted structures for horizontal
diaphragms is rather common in luxury buildings as well in minor ones
in the Aquilan centers.
In the latter, the vaults are generally in stone with thickness varying
between 20 and 30 cm, while in the city of L’Aquila, there were predominantly brick head vaults (12 cm). Barrel vaults create the most comRESEARCH - L’Aquila Earthquake
Fig. 17- Paganica: collapse of two thinly laid brick vaults (horizontal diaphragm on the first floor and a false ceiling on the
second floor).
these vaults collapsed due to the fall of the roof covering structures or
single elements placed above them should be excluded.
2.2.4 Out of plane and in plane response in masonry walls
The topic of masonry quality in the Aquilan territory is the center of
much debate, and not only technical, due to its relevance for interpreting damage but also and most importantly in the choices that must be
made for reconstruction. It is the author’s opinion that judgment on the
quality of masonry cannot be expressed in a univocal manner as of yet.
However, one can certainly attest that most of the building patrimony
was built based on earthquake knowledge, in particular for the systematic use of clamping and tie rods between walls (besides the previously mentioned use in the connections of non-thrusting roof coverings).
The organization of construction systems and the realization of masonry
techniques in many cases calls for the presence of tie rods put in place
while the wall is raised. The construction of such tie rods is easily visible today by the mass of partial collapse dating back to the reconstruction after the earthquake in 1703, when the technique now used for over
two centuries was first put into use. The clamping consists in the placement of a wooden element built inside the wall and connected at the
extremities by an iron plate and nails (Figure 18) and then anchored
externally to the corner by way of a small tie rod (Figure 19). It is very
efficient as long as the wooden element is not placed under too much
tension causing weakness nears the nails.
Obviously, the necessity of balancing costs and the availability of materials called for the elaboration of variations which were not always efficient (Figure 20).
More often than not, the progressive deterioration of these elements was
in correspondence to connections between the wood and the iron plate,
especially in abandoned or poorly maintained buildings.
Generally speaking, the limited presence of out of plane displacement
Fig. 18- San Demetrio: metallic tie rod with a nail connecting the rod to the wooden beam.
Fig. 19- L’Aquila: an elegant 17th century tie rod connects to a wooden beam placed within the masonry thickness (note the
adjacent strengthening tie rod, inserted after construction on the internal side of the wall).
Fig. 20- Villa Sant’Angelo: a poor example of wooden beams placed in the middle of the wall thickness without external
within the vast array of damage found can be attributed to the systematic use of tie rods in building walls.
Nevertheless, it was possible to encounter certain types of out of plane
displacement after this earthquake. Figures 21 and 22 show cracks
which caused detachment from the façade. Both were caused by poor
clamping at the corners and involved of a portion of the orthogonal
walls. In both cases, the portion of the isolated façade was limited to the
height of the horizontal diaphragm.
Figure 23 illustrates a case of the start of overturning of a wall portion
relative to the two floors laid over the ground floor; here the out of plane
displacement seems to have been caused by a lack of clamping with the
masonry adjacent to the contiguous building and a lack of efficient connections at the level of the second floor.
Also damage ascribed to localized displacement of portions of the
masonry cyma were common, both in the form of fractures (Figure 24)
RESEARCH - L’Aquila Earthquake
Fig. 21- Castelnuovo: detachment of a wall caused by defective clamping to the lateral wall.
Fig. 24- San Demetrio: localized fractures above the lintel of an opening.
Fig. 22- Paganica: overturning which involved a portion of the orthogonal wall.
Fig. 25- Paganica: collapse of a portion of a wall over the lintel.
Fig. 23- Tempera: overturning of the top portion of an external wall.
as well as the collapse of limited portions due to cracking near openings
(Figure 25).
In plane damage to walls is present in all of its characteristic variants.
In fact the form and placement of the shearing varied according to the
efficiency of the connections between the elements it was built with and
the quality of the masonry pattern.
Figures 26 and 27 illustrate two cases; in the first, the inclining fractures cross the external wall, traveling indifferently to the horizontal
diaphragms; while in the second, the fracturing is concentrated in the
masonry piers between the openings of the first floor.
These situations are found wherever there was a systematic presence of
clamping or for more recent buildings the presence of stringcourses.
Regarding the validity of anti-seismic regulations, which had been
adopted in most of these buildings, one can only say that in cases where
the workmen had the technical capacity to follow the rules of good con-
RESEARCH - L’Aquila Earthquake
the necessary clamping to the pre-existing masonry; a defect which in
fact greatly limited its seismic resistance.
As previously mentioned, the traditional connecting systems between
converging masonry walls are the most effective: both those put into
place at the time of construction (clamping at the corners, anchored
wooden beams) as well as later strengthening interventions or reparations (tie rods placed in walls). The limited presence of collapse due to
overturning of the façade (compared to damage caused by major in
plane wall displacement) leads one to believe that these systematic connecting measures played an essential role in conservation (Figure 29).
In this sense, the diffuse presence of shear fractures in masonry walls
should not be interpreted as a demonstration of poor masonry quality,
Fig. 26- Villa Sant’Angelo: shearing which crosses the entire external wall.
Fig. 27- Sant’Eusanio Forconese: shear cracks limited to the masonry piers.
Fig. 28- Castelnuovo: masonry buttress built without clamping to the pre-existing wall.
struction and utilized these norms, such measures functioned to limit or
modify damage mechanisms within the limits of their efficacy. Instead,
Figure 28 shows the frequent case of a buttress added to a wall without
Fig. 29- L’Aquila: a building where an overturning mechanism initiated but was effectively contrasted by the connections at
the summit thereby causing in plane resistance behavior of the walls.
but as a positive consequence of the anti-seismic measures which produced a box-like behavior, impeding out of plane mechanisms. In a
well-built stone wall, the formation of even deep shearing following
such a violent earthquake is inevitable and allows the activation of
significant dissipative capacities, limiting the risk of collapse.
Over all, much of the damage observed can certainly be attributed to
insufficient quality of the masonry pattern, whether in ruinous collapse
or the loss of only the external façade. In general, the root of the problem is a lack of building quality: bad equipment for the façade, poor
quality cement (and its excessive proportion to stone elements), absence of transverse connections between walls. Such cases are commonly
caused minor damage to adjacent buildings.
Figure 30 illustrates the case of a building where the entire upper floor
collapsed. One can observe how the seismic activity disarranged the
walls to such a point that it crumbled to the ground in rubble notwithstanding the light roof covering (and therefore not able to induce the
shearing particularly prevalent in upper floors). On the contrary, it
RESEARCH - L’Aquila Earthquake
2.2.5 Localized damage
Always on the subject of masonry walls, localized damage was found to
structural and nonstructural elements, justifiable in a situation where
the force of the seism is extremely violent. In fact, even in cases where
Fig. 30- Castelnuovo: total collapse of the top floor due to defective masonry quality.
should be noted that the external wall of the building to the left was left
In Figure 31, the very common case of the collapse of only the external
wall is shown. In these cases, the modality of damage declared its
cause: the lack of transverse compactness of the masonry panels. Small
or badly placed stone elements, lack of regularity in the pattern and
clamping between the facing sections rendered the walls vulnerable to
horizontal displacement and set off autonomous behavior in both walls
that then very often led to collapse of this kind.
The collapse of a single wall occurred more frequently where r.c. string-
Fig. 32- Casentino: detachment of hewn stone elements around an opening.
Fig. 33- L’Aquila: localized collapse around the borders of the openings.
Fig. 31- Villa Sant’Angelo: collapse caused by the lack of transverse clamping in the masonry mesh.
courses were found at the top of a building, in particular when the entire roof covering was heavy and rigid. The detachment of the wall at the
level of the stringcourse is linked to the local increase of compression
which originates in out of plane flexion from the condition and blocking
vertical displacement (caused by the flexional rigidity of the roof covering).
RESEARCH - L’Aquila Earthquake
Fig. 34- Vallecupa: a chimney exposed by the seism.
Fig. 35- San Pio alle Camere: localized collapse of the top portion due to scarce thickness.
the masonry buildings behaved well, it was natural that there was some
small specific vulnerability.
A frequent case observed was the constructive vulnerability of hewn
stone above doorways and windows. The hewn stone elements used to
border openings were rarely clamped to the masonry walls within the
wall thickness which inevitably led to their detachment (Figures 32 and
33). This glaring omission to the attention given to seismic vulnerability in local anti-seismic construction deserves to be examined more
closely. The problem was probably overlooked considering that such
elements do not determine the stability of the building as a whole.
Further damage was found in consequence to local irregularities of
various types, such as continuity solutions which weaken masonry walls
locally, or the presence of wide parameter walls with no structural function for the building.
In the case shown in Figure 34, one can observe the collapse of the closure of a chimney along the external wall. Figure 35 shows the collapse
of a portion of the top of a wall characterized by scarce thickness, since
it did not bear the load of the roof and was conceived as a simple light
3.1 L’Aquila reinforced concrete building stock
ISTAT 2001 data, representing the official source for information about
building stock in Italy, indicate in L’Aquila city a 24% of reinforced
concrete structures, 68% of masonry structures and an 8% of structures whose typology is not identified (Figure 36a). Data that identify the
age of construction of the buildings (Figure 36b) indicate that 55% of
the whole patrimony was realized after 1945. Considering low incidence of RC structures on the total it is possible to infer that after 1945 new
masonry structures were still realized and that RC structures number
increased gradually over years.
From distribution of storeys number (Figure 36c) it can be observed that
only 5% of the buildings have more than three storeys. Assuming that
all buildings with more than three storeys are RC structures, it is still
possible to infer that the great majority of L’Aquila RC buildings have
no more than three stories.
Assuming an interstorey height that can vary between 3.0 and 3.5
meters, period approximate formulation given by the Italian Code,
referred to RC frame structures, gives for three storeys buildings a fundamental period equal to 0.4 seconds.
In the following section, when comparing different Italian Code spectral
shapes, it will be possible to focus on period value ranges minor or
equal to 0.4 seconds, thus leading to a comparison between constant
acceleration parts of the spectra.
3.2 Structural and non structural damages
In this section main structural and nonstructural damages to RC structures after L’Aquila earthquake are presented. Figuregraphic documentation was produced in the following few days after the 6th April 2009
mainshock. Generally speaking, damages to structural elements are not
so frequent and they seldom involve the whole structural system, on the
other hand damages to nonstructural elements such as internal and
external infill panels involved the main part of RC structures in
3.2.1 Columns and walls
Main structural damages involving RC columns can be easily interpreted as failures caused by mechanisms which capacity design principles
can avoid or at least limit.
It is worth to observe that during an earthquake columns put up with
Fig. 36- 2001 census ISTAT data for L’Aquila town: (a) building typology, (b) age
of construction, (c) storey number
RESEARCH - L’Aquila Earthquake
high flexural and shear demand. Maximum flexural demand joined with
axial force produced by gravity loads and seismic loads are located at
the end of the element; in these zones (critical regions) rotational ductility demand can concentrate. Therefore, it is necessary to provide an
adequate rotational capacity and to avoid buckling phenomena on compressed longitudinal reinforcements.
Actual Italian seismic code provides prescriptions aimed at increasing
section rotational capacity: upper limit on longitudinal reinforcement
percentage, fixed the flexural resistance of the section, leads to a higher
ultimate section curvature; a proper spacing between hoops and crosstie presence give, by a more efficient confinement action on concrete,
an additional increase in section deformation capacity. Additionally, a
proper hoop spacing avoids buckling phenomena in longitudinal
reinforcement or at least fixes an acceptable upper bound limit for
which this phenomena occur.
On the other hand, prescriptions and structural detailing presented
above are typical of modern design concepts that in Italy appeared for
the first time in 1997 with explicative document to the 1996 code but
were adequately ruled only in 2003 with OPCM 3274, and finally officially adopted in 2008 by DM 14/01/2008 and its subsequent explicative document.
So, according to codes that were in force before 1997, it is possible to
find RC columns with longitudinal reinforcement percentage that
exceeds 4% limit, section dimensions not conforming to actual limitations, hoops with a not sufficiently thick space (15-20 cm) and closed
with 90° hooks.
Figure 37(a) presents a corner column of a RC building in L’Aquila
historical centre, probably realized between 1950 and 1960, where
damages occurred at the bottom end section of the element. Presence of
smooth bars and small diameter of the hoops (6 mm) closed with 90°
hooks can be observed, but most significantly the absence of any transversal reinforcement in the first 30.40 cm of the elements immediately
adjacent to the beam-column joint region.
Fig. 37- Column with smooth bars and poor transversal reinforcement (a); damage to a column due to axial - bending
interaction (b).
RESEARCH - L’Aquila Earthquake
Figure 37(b) reports a circular column belonging to a building in the
residential zone of Pettino, realized during ’80s, which shows a typical
damage due to axial force and bending moment; concrete cover was
spalled due to high compression strain and longitudinal bar buckling.
In this case too hoop spacing is not thick enough but probably designed
in perfect accordance with code prescriptions at the age of construction.
In analogy, shear demand can produce brittle failures with an outstanding dissipative capacity reduction of the column. Minimizing shear
resistance mechanism to transversal reinforcement spacing only can be
not exhaustive. In modern design rules shear design has to involve
capacity design principles such as fixing a proper hierarchy between
brittle and ductile failure mechanisms, that is shear and bending behavior of the element.
In order to prevent brittle failures during post-elastic behavior, shear
demand has to be evaluated based on maximum flexural demand of the
element. When this criterion is applied to column, it consists in a rotational equilibrium, obtaining shear demand by the ratio between the
sum of the bending moments at the end sections and the total height of
the column. It is possible to prevent a brittle failure occurrence
applying an amplifying coefficient to the shear demand. When shear
demand is evaluated, possible local interaction with adjacent infills is
considered; in fact when masonry panels do not completely fill the RC
bay, evaluation of shear demand is based on the height of the column
subtracting infill panel height.
Adopting proper capacity models it is possible to take into account concrete degrading resistance mechanisms due to ductility cyclic demand.
These prescriptions are provided by the Italian code since 2003, before OPCM came into force shear design of columns used to be made
assuming shear demand from linear analyses; this procedure can easily
lead to shear capacity underdesigned respect to flexural capacity; so no
control of the failure mechanism can be applied and a priori the failure mechanism can be either brittle or ductile.
All above considerations can be confirmed by column damages reporb
Fig. 38- Shear failures of rectangular (a) and circular (b) columns.
squat column is different respect to the typical behavior of a slender
element. This is why, if local interaction between the column and the
concrete wall is not taken into account, a premature brittle failure due
to concrete excessive compression can more often occur.
Even columns belonging to staircases can show brittle failures. Most
common staircase typologies, generally, are characterized by discontinuity elements in the regular RC frame scheme composed by beams
and columns. In fact, on a side staircase is composed by inclined axis
elements (beam or slab), on the other side squat columns are created by
the intersection of inclined axis element with the column.
Staircase elements considerably contribute to lateral stiffness of the
whole structural system due to axial stiffness of inclined axis elements
and to higher lateral stiffness of squat column. This contribution is simply estimable via linear analyses. On the other hand, staircase elements
are characterized by higher shear demand that can lead to brittle failure mechanisms.
Figure 40 shows a staircase composed by inclined axis beams, in particular the squat column in the corner, due to intersection with the
beam, is characterized by a typical shear failure, due to an unfavorable
shear demand capacity ratio. Poor transversal reinforcement in terms of
spacing and hoop diameter is to be noted.
Shear failures characterized reinforced concrete wall performances too.
ted in Figure 38.
Considering the rectangular column in Figure 38(a), whose section is
probably (30x100) cm2, belonging to a building realized in 80’s, a shear
failure involving top end section is evident. Transversal reinforcement
has a hoop spacing approximately equal to 15-20 cm, that is completely
underdesigned respect to column section dimension and consequently
respect to the section inertia, leading to a premature shear failure of the
element. The brittle failure mechanism is to be noted, underlined by the
crushing of the concrete within the reinforcements. Third and fourth
hoops from the top end of the element that are completely opened.
Figure 38(b) shows shear failure of a circular column characterized by
a 30 cm diameter; in this case too it is possible to detect the hoop spacing not thick enough, that leads to diagonal cracking typical of shear
mechanisms and to instability of longitudinal bars in the column.
In order to highlight the non secondary role played by column – infill
interaction in determining shear failures in the elements, Figure 39 presents some damages to columns. In particular in Figure 39(a) it is possible to recognize the brittle failure in the column due to the local interaction with the concrete infill that partially covers the bay frame getting to 1/3 of the total height of the column. Partial infilling effectively
interacting with the column reduces the effective height of the element,
producing a higher shear demand that exceeds column shear capacity.
This kind of phenomenon involves all of the columns that interact with
the concrete partial infilling.
Figure 39(b) shows an example of buildings with a partly below ground
level. This basement level is characterized by walls, often realized with
concrete, aimed at a retaining function of the adjacent embankment;
concrete wall height is limited respect to column height to allow the realization of windows. This structural solution leads to a strict reduction
of column effective height with a consequent increase in shear demand.
Moreover shear span decreasing of the element can modify shear span
ratio up to a modification to a squat column behavior. This situation is
not of secondary importance, because shear resistant mechanism of a
Fig. 39- Shear failures of column adjacent to partial infilling panels (a), shear failures of squat column adjacent to basement
level concrete walls (b).
Fig. 40- Shear failure in squat columns of the staircase.
Fig. 41- Failure in reinforced concrete walls.
RESEARCH - L’Aquila Earthquake
As an example, Figure 41 reports damages detected in two reinforced
concrete walls, respectively characterized by two different shape factors; damages consist of a spread diagonal cracking. Low longitudinal
and transversal reinforcement percentage can be observed, especially if
compared with minimum values prescribed by actual design code.
3.2.2 Beam - column joints
Beam-column joints can completely modify structural complex behavior and their failure should be necessarily avoided in a proper seismic
design, considering that these element are characterized by brittle failure mechanisms. In these zones, geometrically very small, demand
from beams and columns is concentrated and concrete panel with longitudinal bar is subjected to high gradients of shear and flexural
demand. Beam-column joints influence structural behavior in terms of
both ductility when concrete cracking and bar sliding occur and resistance when brittle failures occur.
Failure mechanisms of joints are principally governed by shear and
bond mechanisms. In fact force distribution which allows shear and
moment transfer produces diagonal cracking and consequently joint failure due to diagonal compression in the concrete can occur, producing
a reduction in strength and stiffness in the connection. Cyclic degrading
of bond mechanism, on one hand, produces a reduction in bending resistance and in the ductility of the elements meeting in the joint; on the
other hand an increase in lateral deformation of the level is detected.
Therefore, capacity design rules essentially aimed at avoiding brittle
failure mechanisms point to shear failure prevention in joints by means
of design rules and proper reinforcement detailing. In fact if the joint
collapses, a strict limitation results in resistance and deformation capacity of the adjacent structural elements.
Generally speaking, joint design is defined by the condition of a diagonal stress induced by the elements meeting in the joint not exceeding
allowable concrete compressive stress. Furthermore, in order to keep
structural continuity when concrete cracking occurs, a proper transversal reinforcement along the whole element should be provided.
Transversal reinforcement allows to transfer stresses even if concrete
cracking phase has been overtaken, by means of a strut and tie mechanism that can develop if longitudinal, transversal reinforcement and
concrete struts contribute to a truss formation. By these design prescription in the joint a ductile mechanism in beam elements can develops, avoiding a brittle failure in the joint.
Joint design rules appeared in the Italian design prescriptions only in
2003, in fact in 1997 explicative document to 1996 code a transversal
reinforcement in joints at least equal to the hoop spacing in the adjacent columns was simply necessary.
It can be gathered that all structures realized before 1996 are characRESEARCH - L’Aquila Earthquake
terized by beam-column joints without transversal reinforcement. In
this kind of situation, when cracking occurs no truss mechanism can
develop, leading to a strict reduction in strength capacity of the joint.
Considerations presented above are confirmed by damages observed in
RC structures after 6th April 2009 earthquake. Figure 42(a) shows an
external beam-column joint, characterized by an extensive cracking in
concrete belonging to the joint panel. The absence of transversal
reinforcement in the joint, probably because of the vertical component
of the seismic action, led to local buckling of the longitudinal bars that
consequently brings to concrete cover spalling. It is worth to observe
that the absence of a proper transversal reinforcement in the joint conducted to a loss of anchorage in beam longitudinal reinforcement.
Figure 42(b) shows a typical diagonal cracking failure in concrete panel
belonging to an external joint. Crack begins at the intersection between
joint and upper column and ends at the intersection between joint and
lower column producing the loss of monolithic connection. Hoops
absence, in this situation too, leads to a buckling in the external longitudinal bar and it involves lower column not provided of transversal
reinforcement in the first 30-40 cm.
Other remarkable aspect in RC damage observation after L’Aquila
earthquake is a peculiar loss of connection at joint-lower column interface, probably due to a not prepared cast surface that can lead to a shear
friction failure. Generally, there are no code prescriptions, nor Italian or
international, providing a shear friction verification at joint-column
interface, because some execution details, such us preparation of the
cast surface and a proper diffusion of longitudinal reinforcement along
the perimeter of the column section ensure that this failure condition
does not limit or influence design procedure. In fact, main resistant
mechanisms in post-cracking condition are referred to (i) concrete to
concrete interface shear, (ii) dowel action in column longitudinal
reinforcement and (iii) clamping action produced by a local yielding in
longitudinal bar that contributes to transfer main part of the shear
Fig. 42- Joint failure with evident longitudinal bar buckling (a), diagonal cracking failure in concrete joint panel (b).
Fig. 43- Failure mechanisms at joint – column interface surfaces.
Shear friction mechanism is evidently influenced by axial force amount
and roughness (friction coefficient) of the casting surface; a not proper
preparation and control of the casting surface can reduce shear friction
Clamping action – a friction mechanism too, integrating above contributions – is proportional to longitudinal reinforcement amount.
Dowel action, not negligible in post cracking phase, is strictly connected to longitudinal reinforcement percentage but mainly, due to their
peripheral position, to effectiveness of transversal reinforcement in the
zone adjacent to the sliding surface. Hoops, in fact, work as a translational restraint to longitudinal bars involved in the mechanism.
Figure 43(a) reports a failure mechanism due to the loss of continuity at
the intersection between joint and lower column. A poor treatment of
the casting surface can be easily detected and it can lead to a lower concrete to concrete friction coefficient in correspondence with this weaker
surface. Reinforcement probably composed by only four longitudinal
bars and the absence of hoops in the zone strictly reduce dowel action.
In fact, this limitation is substantially due to a low longitudinal reinforcement percentage not properly spread on the section perimeter, joined
with the spalling of the concrete cover that is the only transversal
restraint to horizontal bar sliding. A strong reduction in the axial force
essentially due to a non ordinary strong vertical component of L’Aquila
event has further reduced shear friction capacity. Figure 43(b) shows,
as well as Figure 43(a), a clear separation in concrete at joint-column
interface. Absence of hoops in the joint and axial force reduction due to
vertical seismic action strictly reduce, respectively, dowel action
mechanism and friction mechanism at joint-column interface.
3.2.3 Infills
In the previous section it was emphasized how damage limitation limit
state prescriptions and verifications are essentially aimed at avoiding or
reducing infill damage and most remarkably that this kind of prescription was firstly introduced in the Italian code only in 1996 and better
detailed and completed in 2003 with OPCM 3274.
Therefore, it is reasonable to assume that the main part of RC buildings
in L’Aquila was realized without any deformability control and verification. On the other hand it should be emphasized that, even if a design
procedure according to 1996 or better according to 2008 code had been
employed, involving damage limitation verification, the strong PGA
characterizing L’Aquila event would have equally produced a spread
diffusion of damages to nonstructural elements such as external infills.
As a general rule, infill failure mechanisms can be classified in: (i) horiFig. 44- Infill panel failures: diagonal cracking (a), (b) and corner crushing (c)
RESEARCH - L’Aquila Earthquake
zontal sliding in the central zone of the infill panel, (ii) diagonal
cracking due to tensile stress in the central zone of the infill panel, (iii)
corner crushing in the direct contact application zone.
Figure 44 reports two building facades in which infill panels are characterized by a diagonal cracking mechanism. In the first case, see
Figure 44(a), it is worth to note how cracking diffusion involves infilling
adjacent to window openings and damage is concentrated at the first
levels of the building; in the second case, Figure 44(b), diagonal
cracking is more emphasized by plaster layer because the external layer
of the infill is composed by solid clay bricks. Figure 44(c) shows a typical corner crushing mechanism. Out of plane failure of the infilling
external layer gives the possibility to detect corner crushing mechanism
of the internal layer; other evidence is the crack, that visibly involves
the plaster but probably is deeper, localized at the top of the column
adjacent to the infill panel as a consequence of column-infill local interaction.
The great majority of external infill panels are composed by double
layer infill panel, internal layers are generally realized with clay bricks;
connections between the two layers are realized by the interposition of
brick elements discretely, Figure 45(a), or lined up, Figure 45(b). Not
reliable efficacy of this system should be stressed.
Furthermore, in most of the observed cases, internal infilling layers are
restrained at the four corners of the RC frame while external ones are
constrained only by the upper and lower beam by means of a little pawl.
This executive solution leads to a reduction in the interaction mechanism
between RC frame and external infill panel in both plane and out of
plane seismic forces. In fact, the low efficacy of the restraint applied to
the external panel, coupled with ineffective connections or complete
absence of connections between the two layers, produces a damage
restricted to the external infill panel, which shows an out of plane failure due to seismic action in both directions as it can be detected in
Figure 46.
Windows or door openings represent a discontinuity in the infill panel,
modifying its performance and capacity by a reduction in terms of stiffness contribution and by a modification in the failure mechanism.
Figure 47 shows damages detected in L’Aquila RC buildings characterized by a different opening position in the panel or a different opening
percentage respect to the total area of the bay.
Both local and global interaction effect between infill and RC structure
are not negligible. As it was previously emphasized, local interaction
between infill panel and adjacent column can bring to (i) a reduction in
the effective length of the column, an increase in shear demand and a
consequent brittle failure of the column when the panel partially fills
the frame bay; (ii) to a concentration of shear demand at the end of the
column and to a consequent brittle failure when diagonal compression
RESEARCH - L’Aquila Earthquake
Fig. 45- External infill panel failures and discrete (a) or lined up (b) connections between layers.
Fig. 46- External infill panel failures without connection between layers (a) and with ineffective connection (b).
is applied by the panel to the RC element.
As a global phenomenon, infill-structure interaction increases global
stiffness of the complex system and consequently spectral acceleration
demand, besides it can represent a source of irregularity in plan or in
elevation when a non uniform distribution is present.
3.2.4 Criteria for regularity in plan and in elevation
Regularity criteria in plan or in elevation were introduced in the Italian
code in 1997 by the explicative document to the 1996 code as a qualitative definition. Only in 2003 by means of OPCM 3274 introduction a
Fig. 47- Infill failure mechanisms differing form opening position and percentage.
quantitative definition of regularity in plan and in elevation was furnished.
A compact shape in plan that implies geometrical limitation in the
appendixes, a uniform distribution of resistant systems in plan, slabs
that can be considered as rigid respect to vertical resistant elements can
ensure a regular structure in plan.
Conversely, a verification in mass and stiffness distribution, verifying
that storey lateral strength is not characterized by unexpected reduction, leads to a regular structure in elevation.
In spite of a quantitative definition of regularity criteria, not necessarily
these criteria are enough to avoid other irregularity sources not considered in the design procedure of the RC structure. It is the case of infills
that by means of interaction with the structure can strictly modify stiffness and strength distribution in plan and in elevation.
It is necessary to avoid non uniform distribution of infills or conversely
it is necessary to explicitly take into account infill contribution in the
analysis procedure. For example, Italian seismic code provides some
prescriptions aimed at considering infill irregularity contribution in
plan by an increase of accidental eccentricity value or in elevation by
an increase of the shear demand at the storey characterized by an irregular distribution of the infills.
Some peculiar cases of structural failure after L’Aquila event, mainly
caused by irregularities in plan or in elevation, are reported in Figure
48. Figure 48 shows a view of the structures before the earthquake, so
without any damages, on the left, and after the event where collapse is
evident on the right.
The first structure, Figure 48(a) and 48(b), was placed in the centre of
L’Aquila city (Porta Napoli street); it was characterized by elevation
irregularity due to the non continuity of the seismic resistant scheme
over height, additionally, second level was characterized by an evident
discontinuity in terms of infill distribution, in fact on the left wing of the
building there is a sort of porch. Observing building collapse, a damage concentration at the second level is to be noted, that consequently
produced complete failure of the upper levels.
The two other structures proposed in Figure 48 are both placed in the
residential zone of Pettino (Dante Alighieri street), close to L’Aquila,
and present the same shape in plan similar to a T, so they both can be
considered as irregular in plan. In addition to plan peculiarities, a different distribution of infills at the first level respect to the others, due to
the presence of garages entrances, can be observed. Both buildings placed in this street showed a soft-storey mechanism at the first level that
can be explained by infills irregularities in elevation and presumably by
a local interaction between infills and adjacent columns, leading to a
brittle failure of some columns at the first level.
Fig. 48- Soft storey mechanisms examples in L’Aquila: Porta Napoli street (a), (b), Dante Alighieri street (Pettino) (c), (d), (e)
and (f) before and after collapse occurrence.
One of the main objectives of the Civil Protection in the immediate
post-earthquake of L’Aquila of April 6th, 2009 has been scholastic buildings’ damages assessment as well as the fast repair of the ones with
non structural damages only. These activities have been developed by
a joined work of Function 1 of Emergency Management and Quarter of
Department of Civil Protection (DPC), Consortium ReLUIS, Seismic
Risk Competence Centre of DPC and Public-Works Office of Lazio
Sardegna and Abruzzo.
The structural safety assessment of L’Aquila scholastic buildings started on April 8th, 2009. The in-site inspections have been coordinated
by ReLUIS under the supervision of Function 1 of DPC at Reiss
Romuli; the inspections have been related to both L’Aquila and its provinces scholastic buildings.
The activity involved 62 scholastic buildings of L’Aquila: 53 under the
administration of L’Aquila municipality (6300 students on the total of
about 7000) and 9 of province (4000 students on the total of about
5000). A total of 156 structures have been investigated. The results of
structural safety assessment is summarized in Figure 49.
The in-site inspections on scholastic buildings in the L’Aquila provinRESEARCH - L’Aquila Earthquake
and about 21% masonry structures (see Figure 51)
In terms of damages the results of in-situ inspections showed that about
31% of RC framed structures were assessed as A (i.e. no significant
damages), about 43% as B (i.e. no significant damages on structural
members) and about 26% as E (i.e. significant damages on both structural and non-structural members). The structures with a lower level of
damage (i.e. A and B) have been mainly built between ’60s and ’90s
years while the structures recorded as E were mainly built between ’20s
and ’70s years. The RC framed structures mainly showed damages on
non-structural members (i.e. partitions and ceiling); the elementary
school of Paganica is a typical example of such kind of structures (see
Figure 52-53). An example of RC framed structure with significant
damages on structural members is the school “Celestino V”, see Figure
54-55. The RC shear wall-type structures were assessed only as A
(27%) or B (73%); no significant damages on structural members were
found at all. The masonry structures, mainly built before ’60s years,
were assessed as: 30% A, 24% B, and 46% E. Figure 56-57 show some
significant damages on masonry members of the elementary school
Fig. 49- Structural safety assessment on L’Aquila scholastic buildings.
Fig. 50. Structural safety assessment on scholastic buildings of L’Aquila provinces.
ces were performed in 64 different municipalities on 224 buildings for
a total of 309 structures; the results of such activity is summarized in
Figure 50.
The scholastic buildings of L’Aquila are mainly reinforced concrete
(RC) or masonry structures; in particular, about 66% are RC structures
(56% RC framed structures and 10% RC shear wall-type structures),
Fig. 52- Damages on partitions at elementary school of Paganica.
Fig. 51- Structural safety assessment results on scholastic buildings of L’Aquila: a) A; b) B; c) E.
RESEARCH - L’Aquila Earthquake
Fig. 53- Cracks and plaster spalling on partitions of elementary school of Paganica.
Fig. 56- Roof collapse of elementary school “S. Elia”.
Fig. 54- Column crack at elementary school “Celestino V”.
Fig. 57- Diagonal cracks on masonry panels of elementary school “S. Elia”.
Fig. 55- Damages on ceiling at elementary school “Celestino V”.
In the immediate post-earthquake several teams with members from
ReLUIS, Seismic Risk Competence Centre of DPC and Public-Works
Office of Lazio Sardegna and Abruzzo were involved in a further stage
of in-situ inspections on schools in order to investigate the repair possibility before the new scholastic year official opening (foreseen in
September). In some cases several destructive and non-destructive tests
were performed in order to investigate on the materials mechanical properties. As a result of this second round of in-situ inspections, a cost
estimate to fully repair these schools was also performed. The repair
interventions were planned on schools assessed as A and B; the repair
strategy and the interventions design were provided by engineers of
municipality or province under the supervision of ReLUIS and PublicWorks Office. The Public-Works Office also managed the bids for works
execution (see Figure 58).
RESEARCH - L’Aquila Earthquake
Fig. 58- Table of works at elementary school of Torrione.
Fig. 61. Intervention on partition to avoid the overturning.
Fig. 59- Strengthening of joints on elementary school of Paganica.
Fig. 62- Investigation on masonry corner of elementary school “Villa Grande” of Tornimparte.
Fig. 60- Strengthening of joints on elementary school of Torrione.
According to Ordinances 3789 and 3790 and commentaries the works
involved not only the repair of non-structural members but also local
strengthening interventions on structural members (i.e. strengthening of
front and corner joints of RC structures (Figure 59-60), insertion on
masonry members of chains and braces) and non structural members
(interventions on partitions in order to avoid their overturning, to connect their internal and external faces, and application of zinc coated
steel grids on partitions), Figure 61. The execution of repair and
strengthening works has been developed together with the execution of
materials non-destructive tests (rebound and sonic tests, tests on concrete cores and steels specimens, (Figure 62) as well as tests by using
flat jacks on masonry structures). Finally, load tests were performed on
slabs and flights.
The analysis of several cases of study allowed to optimize a series of
repair or local strengthening intervention on existing RC or masonry
RESEARCH - L’Aquila Earthquake
structures; a detailed description of standard repair interventions to
restore damaged buildings due to seismic actions has been reported in
a document developed in collaboration between DPC and ReLUIS,
“Guidelines for Repair and Local Strengthening of Structural and Non
Structural Members”.
The synergy between DPC, ReLUIS, Public-Works Office as well as
municipality and provinces allowed scholastic buildings assessed as A
to be opened with urgency in order to regularly perform the exams at the
end of the scholastic year. Further, the whole stock of scholastic buildings assessed as B has been opened in the period between September
21st and October 5th.
The recent earthquake of 6th April 2009 significantly hit the city of
L’Aquila and its surroundings both for the serious number of casualties
and for the damage suffered by residential and important structures.
Among the latter, one of the most important is surely the San Salvatore
Hospital of Coppito, the crucial point of the hospital system in the area
of L’Aquila, which was completely evacuated during the emergency due
the damage at various floors of the buildings.
A thorough analysis of the hospital facilities has shown besides the
essentially non-structural damage also some aspects of the construction
of the hospital complex, which might be crucial for the seismic response to future shocks and therefore open more critical damage scenarios.
Other smaller hospitals located in neighbouring urban areas showed
minor damage, allowing partial absorption of emergencies, thus reducing the enormous overload on the field hospital set up close to the San
Salvatore Hospital. Recent valuations, in fact, have shown that more
than 1500 injured received assistance during the days after the emergency, which confirms the tremendous impact on the hospital system.
The hospital structure has been designed in 1966 and took about 30
years to build entirely. Such an information is particularly important if
related to developments in domestic seismic regulation. In fact, the first
Italian seismic law, considered a forerunner of the most modern ones,
was n. 64 of 2/2/1974, issued after the 1974 Ancona earthquake.
Previously, references to seismic law came from Royal Decrees (e.g.
Regio Decreto Legge n. 2105 of 22/11/1937), while during the Republic era there was Act n. 1684 of 1962 that followed the Campania
earthquake of 21/08/1962, and was later completed by Act. n.1224 of
5/11/1964, and by the Act n. 6090, 1969 issued after the Belice earthquake. However, such references were oriented towards defining heights,
thicknesses, executive methods and quality of materials rather than calculation methods and design criteria.
After the 6th April earthquake, only 3 buildings out of 15 in the San
Fig. 63- Evidence of the reinforced concrete frame structure through the damaged partitions: Building 9 (a); Building 10 (b).
Salvatore complex suffered considerable structural damage. This was
limited to small areas and primarily was due to evident issues that will
be described in detail later in this work. There was slight and relatively
limited non-structural damage and significant non-structural damage in
only a few buildings. In these latter inner partitions significantly helped
the lateral resistance by dissipating the earthquake’s energy and suffering critical damages (Figure 63).
The basement and semi-basement part of the complex, mostly made up
of reinforced concrete walls, showed a stiff box response without significant damage. Finally, no evident damage was observed in the foundation structures.
From the point of view of the human safety, the most widespread and
relevant non-structural damage was that of exterior façade bricks, covering the entire surface of all the buildings. Such a coating, not linked to
the interior walls, in many cases was partially or totally detached. No
significant damage was observed on the equipment and internal mechanical devices inside.
5.1 Description of the structures of the San Salvatore Hospital
In terms of typology, the San Salvatore Hospital complex consists of a
series of reinforced concrete frame structures, with interior and exterior
masonry walls, built from the mid ‘70s on, and put into service in the
second half of the ’90s. Some of the buildings of the complex are not
hospital property.
The buildings differ in typology, materials and heterogeneous construction details depending on the different age of construction. A covered
walkway connects the various blocks on four floors, two above ground
and two underground.
There are several building typologies: L-shaped of 2 or 3 storeys, tower
blocks of 3 or 4 floors, in-line buildings of 2 or 3 storeys and some
ground-level buildings. Most of them have one or two basement floors.
The approximate date of construction and the main function of the
various blocks, identified on the basis of the numbering given in Figure
64, was provided by the Technical Department of the hospital and is
shown in Table 1.
RESEARCH - L’Aquila Earthquake
Table 1 - Buildings of the San Salvatore Hospital complex, Coppito,
Construction age
Building 1
Thermal Power station
and refectory
Building 2
Analysis laboratories
Building 3
Diagnostics and radiotherapy 1976/77
Building 9
Emergency room
Building 10
Pharmacy and operatory rooms 1978/79
Building L1
Building L2
Obstetrics and gynaecology
Building 6
Building L3
Building L4
Infectious Diseases
Building L5
Delta 6
Delta 7
Medical Delta
Delta 8
Surgical Delta
In all the buildings of the complex there are also several structural joints
of sizes and characteristics not appropriate to shock induced movements. The resulting local pounding between adjacent bodies caused
localised damage, somewhere particularly evident. Figure 66, for example, shows the damage to a structural joint ending on the top of a column,
which have caused an abnormal concentration of pressure on the joint.
Figure 67 shows the cracking continuing in the ceiling from the damaged joint in the wall, found in a connecting walkway (Building 2).
Some local damage have been caused because of improper construction
5.2 Usability surveys and structural response
Fig. 65- Damage to the coating of Building 2 (a), Building 9 (b) and Building L3 (c).
The buildings of the hospital complex have been repeatedly checked
because of a succession of significant after-shocks related to the so-called seismic swarm, and were grouped into categories depending on the
assessed damage.
All the buildings have an exposed brick wall covering, which is not
enough or not at all connected with the infill panels and the reinforced
concrete frame structure. These are therefore critical for human safety
since they can be dangerous and might fall onto the walkways (Figure
65). Another type of frequent and potentially dangerous non-structural
damage is the complete detachment of the coating of many of the
ground floor tiled walls.
Fig. 66- Improper structural joint ending on the top of a column.
Fig. 64- Site plan with different blocks (the groups of buildings identify the
classification of usability as described in section 4).
(Gruppo/Group; Edificio/Building)
RESEARCH - L’Aquila Earthquake
Fig. 67- Damaged structural joint on wall, extending in ceiling.
shortly, and therefore can be usable by means of simple short term
countermeasures, among which: removing falling cladding, plaster and
detached coatings, making safe damaged false ceilings, repairing light
damage to claddings and partitions, checking all the fixings hanging
from cladding and false ceilings, in order to evaluate the possible detachment risk, creating localized barriers to protect walkways, and removing unsafe portions of the outer coating. Buildings L1, L2 and L5
belong to this category, which show widespread light non-structural
damage, particularly at the ground level. These three buildings will be
the first to be reopened at the end of May 2009, less than two months
after the main shock.
Group 3: partially usable buildings
Each of the Buildings 2 and 3 is composed of two different blocks separated by a structural joint, which show different levels of damage. In
particular Buildings 2A and 3A, which face Building Delta 7 (Figure
69), are not usable, whereas light non-structural damage is common to
the rest of Buildings 2B and 3B (Figure 69).
Unlike group 1, in this group of buildings the unusable part features
high structural risk, whereas the less damaged part requires modest
intervention before to be used again.
Fig. 68- Examples of improper transversal reinforcement and insufficient concrete cover in Building 2A and in the external
connecting walkways.
detailing, at least according to current seismic design criteria, such as
the not always appropriate confinement of the structural elements, the
insufficient concrete cover (Figure 68), the presence of columns made
squat by the infill masonry walls.
Fig. 69- Buildings 2A/3A and 2B/3B (Edificio/Building).
Group 1: usable buildings
Regarding the basement, 6 metres underground, where the thermal
power station is located, Building 1 shows moderate non-structural
damage, particularly in the offices at the lowest floor. The portion housing the power plant shows a moderate damage in some beams, mainly
as a result of the relative movement of the Gerber half joint, and it is
therefore usable.
The rest of the building, however, shows greater damage in the upper
storeys, even if it does not constitute a danger to the floor below, and it
is therefore to be considered unusable.
Group 2: buildings that can be made usable with short term countermeasures
Some buildings with slight non-structural damage can be reopened
Fig. 70- Damage of the stairs of Building 2 (a) and Building Delta 8 (b).
Buildings 2B and 3B can be made usable with short term countermeasures, like those described in the previous paragraph. At the last flight
of the inner stairs between portions 2A and 2B, mild localized structural damage was detected, probably due to concentrated rotation of the
RESEARCH - L’Aquila Earthquake
Fig. 71- Structural damage (plastic hinges and shear failures) at the top of columns (ground floor of Building 2A) caused by insufficient transverse reinforcement.
Fig. 72- X cracks on the external coating of Building 2A.
Fig. 73-. Important damage to partitions in Building 9.
RESEARCH - L’Aquila Earthquake
Fig. 74- Structural damage caused by the presence of belt windows, which make
the columns squat in Building 10.
knee beams (Figure 70). In this case limited restoration intervention
can be provided, such as repairing cracks with resin or anti-shrinkage
mortar, and later if necessary a seismic retrofit can be designed using
fiber-reinforced materials.
Blocks 2A and 3A are unusable because of the considerable non-structural damage and the structural risk caused by the failure of some
columns of the covered walkway over the driveway to the Emergency
Room, facing the Building Delta 7. This is the first area of significant
structural damage. As shown in Figure 71, at the top of the columns it
is possible to recognize clear plastic hinges, shear failures, spalling of
the concrete cover with the consequent instability of the longitudinal
reinforcement. All these effects are due to the scarcity or absence of
transversal reinforcement, i.e. of confinement. This has led to a near
state of collapse, requiring urgent propping intervention. In the same
area there is widespread damage to infill panels, with the formation of
X-cracks (Figure 72) typical of shear failure. Although such failures
may affect only infill panels, potential shear failure to the structural elements under the damaged coating cannot be excluded, and therefore
they must be checked.
Group 4: buildings requiring more extensive non-structural measures
A number of buildings can be made usable only after significant local
demolition and reconstruction of the most damaged partitions and removal and restoration of all unsafe or detached parts. The restoration of conditions of temporary usability in this case requires more time, and therefore cannot be included among short term countermeasures.
Buildings 9, L3 and L4, in this group, have suffered moderate to severe
non-structural damage to many of the ground floor partitions (Figure 73)
and a widespread light non-structural damage at the upper storeys.
On the more damaged part, the usability restoration requires demolishing
and rebuilding some infill walls, and propping a stair in Building 9.
Group 5: unusable buildings
The last group of buildings is classified as unusable, due to the significant structural damage or because they are close to dangerous buil114
dings, or because they feature a high percentage of severely damaged
partitions compared to the volume of the building.
Building 10 has severe structural damage at the ground floor, because
of the shear failure of all the columns on the side in front of the church.
In this part the columns have been made squat by the infill panels interrupted by the belt windows along the north-east side (Figure 74). It is
clear that for most of these columns the bearing capacity is seriously
compromised. The building also has widespread moderate to severe
non-structural damage, particularly at the ground level.
To ensure this building will not collapse, short-term countermeasures
are arranged, by means of a series of cement block infill panels between
the columns of the external porch and by propping the beam supported
by the damaged columns.
In Buildings Delta 7 and Delta 8 there is widespread structural damage more important at the lower levels. The damage to partitions is nonstructural but widespread and of moderate to severe intensity. The only
light structural damage is found in one of the stairs.
The two buildings are considered unusable, due to the significant extension of damage, together with the structural irregularity and the proximity
of buildings with structural problems (Building Delta 8 is next to Building
10 and Building Delta 7 is adjacent to Buildings 2A and 3A).
5.3 Usability of walkways and connections in basements
The basements of all the inspected buildings, where accessible, have no
Fig. 75- Connecting walkways: hollow clay block collapse of the intrados of the basement at 3 metres below ground.
RESEARCH - L’Aquila Earthquake
significant damage. There are rare and limited examples of hollow clay
block collapse in the ceiling of basements at 6 and 3 metres below
ground (Figure 75).
At the level of 6 metres below the ground there are unused spaces,
accessible only by technical personnel, without significant damage,
except for localized seepage and settlement/shrinkage cracks.
At the level of 3 metres below the ground, in the underground passage
connecting the buildings, the damage is predominantly non-structural
and not very extensive. It mainly consists of the damage to existing
joints and the formation of ‘natural’ joints after the event, with limited
plaster and coatings detachments, and some hollow clay blocks collapse.
In such basement levels the damage is more significant towards Building 1, where some localized modest damage is present.
At the first floor level (3 metres above ground) there is damage to existing joints, formation of cracks, primarily to the completion of incomplete joints, and the subsequent detachment, sometimes only partial, of
plaster, false ceilings and coatings.
At the ground level the most critical situation is found since the connection walkways between the various buildings are not protected,
sometimes there is just a covered porch. The risk of falling of loose parts
of partially detached coating or new portions of coating falling off, even
after minor shocks, makes the need to remove partially detached or
already collapsed coating urgent, to protect all the paths adjacent to the
buildings (when there is no covered porch) from falling objects and to
close off the riskiest paths. At this level there is some structural damage: a column behind Building 9 stressed by a structural joint, and
another column on the corner of Building 2 with a plastic hinge at the
6.1 Building-like industrial structures
Industrial buildings were built for many years as an assemblage of precast reinforced concrete elements. The April 2009 L’Aquila earthquake
has struck, for the first time in Italy, industrial structures on a large
scale. In fact, the Irpinia 1980 earthquake hit an area with few industrial sites; and similarly happened in the Umbria and Molise earthquakes, which moreover were felt within a limited area. On the other
hand, the Friuli 1976 earthquake damaged industrial structures, but
they were designed with no regard with respect to the seismic action; if
any design rule was used, this however belonged to inadequate seismic
codes. L’Aquila and its surroundings are instead undergoing a rather
strong industrial development. Precast reinforced concrete buildings,
RESEARCH - L’Aquila Earthquake
with a column-beam structure, are the usual type in the industrial areas
in Pile, Bazzano, Monticchio and Ocre (AQ). Buildings have generally
one storey; two storeys are seldom observed, moreover only on a more
limited area than the first storey. Beam supports on the columns may be
of the saddle or bracket type. Deck beams are usually transversally aligned (with respect to the longer side of the structure plan) and they are
I-shaped, with variable depth; longitudinal alignment and inversed Tshapes are less commonly observed. In the latter case, the shape is used
to support the tiles. The roof is generally built with tiles -shaped; less
often U-shapes are observed. Skylights are sometimes present. -tiles
are also used for the intermediate deck, when there is one. External partitions are either made with bricks, or with precast reinforced concrete
shells, with no stiffeners.
Structural nodes are those typical of the Italian constructions: beaker
footing for the foundation to column nodes, simple support for the beam
to column nodes (with neoprene bearings and steel pins). Tiles are
generally directly resting on the beams, with neither horizontal restraint
nor neoprene bearings. Pin connections are seldom present. Tiles may
be connected each other with the upper reinforced concrete layer, or
simply linked via steel restrainers (partly poured within the tiles concrete, partly welded with the next tile restrainer). Partition panels are
either supported by the eaves beam or by the column, via links of many
types. They are also sometimes supported by the deck tiles. A typical
shell-eaves beam connection is via a steel plate partially put within the
concrete shell during pouring. A bolt connects the plate and a steel
angle, which is restrained to the eaves beam edge. The shell to column
connection is often built with a steel plate within the column and a
bayonet with bushing linked to the shell, via a long bolt. This technology is used also to connect the partitions to the deck tiles; the steel
angles and bolts connection is less frequent. It is worth to note that
structural shell buildings are much less common than beam-column
buildings. A few precast industrial buildings were under completion on
April 6th, 2009, when the earthquake struck; so it was possible to verify
the seismic behaviour of these structures under variable degrees of
6.2 Damage and seismic performance analysis
The response of the structural elements of the industrial buildings to the
April the 6th 2009 earthquake was generally in accordance with their
design level: no column collapsed, even though, in many cases a plastic
hinge was observed, due to the high intensity of the seismic action
(Figure 76). In some cases such plastic hinge was not observed at the
column base, i.e. at the column-foundation joint, but even one meter
above, where the longitudinal reinforcement decreases. Furthermore no
Fig. 76- Plastic hinge in columns of industrial buildings: (a) FIAT garage at Pile; (b) building used for bovine-breeding at Fossa.
Fig. 78- Collapse of beams due to loss support of a building at Fossa used for bovine-breeding.
ve the collapse of the beam due to support loss at the side without joint
bar, caused by too large displacements, and the pounding between the
beam and the column top fork (Figure 78).
The phenomenon of the joint bar cover splitting can be also noted at the
intermediate level of some two-storey precast buildings, where, as
already written, the beam-column joint is on corbel. The same phenomenon has also characterised the collapse of some tiles. In this case,
indeed, even where the joint had been fastened by a steel bar, the little
thickness of the bar cover of the beam, also characterised by the lack of
stirrups, collapsed, causing the tile support loss (Figure 79a).
Obviously, such support loss easily happened where the tile-beam connection was not fixed and/or there was no connection between tiles; particularly unlucky situations were characterised by buildings in phase of
assembly, where the floor slab, joining the tiles, was not made yet.
Fig. 77- Effects of pounding between the cover tiles and the beam.
plastic hinge was observed in beams or tiles due to the increment of the
vertical action. However, the damage of the precast industrial buildings
should be well analysed; indeed, it was characterised by collapse of
parts of the buildings, which, if the mainshock had happened during the
working time instead of at 3 a.m., it would have caused victims.
The static scheme of such structures is characterised by large deformability; consequently, the most of the observed damages of structural elements (made by reinforced concrete) depend on the relative displacements between the elements. Indeed, many cases of pounding between
elements of the same structure were observed. Furthermore, pounding
between adjacent buildings was frequent, in the case of both precast
and cast in situ structures, due to the insufficiency of separation joints.
In Figure 77 the pounding between the tiles and the beam of an industrial precast building placed at Bazzano is shown.
Confirming the numerical studies performed in the last years, the connections represented the weak parts in terms of seismic performance of
both old and new precast buildings. Some buildings have shown damages at the beam-column connection: the only observed case of precast
beams collapse was due to the damage of such connection and to the
following support loss. Indeed, as shown by numerical analyses, the
splitting of the joint bar cover happened where the thickness was minimum. In other frames of the same structure, it is also possible to obser116
Fig. 79- (a) Tiles collapse due to cover splitting and support loss of a FIAT garage at Pile (b).
Collapse of perimeter panels due to the breaking of the angle stirrup or
to the bolt head going out from the profile happened in a building used
as material and machine deposit at Bazzano.
However, the most important and spread damages of precast industrial
buildings caused by the April 6th earthquake are those concerning the
elements on the perimeter; indeed, the large damage of such elements,
even though the structural typologies are different, associates precast
buildings to the in situ cast ones. The top connection of the vertical
panels to the side of the gutter beam, made by a profile drowned in the
panel, bolt with nut and angle stirrup, in some cases gave way due to
the angle stirrup breaking and/or to the bolt head going out from the
profile (Figure 79b). This last phenomenon also caused the collapse of
panels connected to columns by a profile drowned in the column and
bayonet with bushing joined to the panel by bolt; some of these last connections, instead, collapsed due to the bayonet breaking at bushing
RESEARCH - L’Aquila Earthquake
Fig. 80. Local failure of the beam support.
position. Some other failures where due to the going out of the whole
profile from the panel where it was drowned. A better seismic response
was shown by panels joined to the structure by angle stirrups and bolts.
In the case of perimeter elements made by bricks, the seismic action
determined their out of plane deformation, in many cases up to the
expulsion of bricks and the consequent partial or total collapse of the
perimeter element. Finally, among the carrying out mistakes, it is
noteworthy, for precast structures, the local failure of the beam support.
In Figure 80, a near collapse condition is shown, caused by the large
column cover due to the fire protection provisions; indeed, due to such
provisions, a volume of concrete without reinforcement works as beam
Fig. 82- Pictures of the sili after the seism (Figure by G. Verderame, ISPRA, F.M. Mazzolani).
6.3 Non-building-like structures: the case study of the Sili Vibac at Bazzano
The sili of the Vibac multinational (a chemical company which produces plastic films), located at Bazzano, close to Onna (Figure 81) represent an exceptional case of damage to steel constructions. They also
represent an emblematic case of damage induced by the earthquake of
April 6th. The sili are used for the storage of polypropylene pearls, and
they were full when earthquake struck. Some sili collapsed, some other
remained standing even though strongly deformed, both locally at some
rings and diffusely (Figure 82).
A more close visual inspection indicated that the collapses occurred for
overturning due to the crushing of the base rings and the hopper.
Moreover, along the sili height, deformations induced by buckling phenomena of the wall panels are apparent. In some cases an effect of
pounding on the adjacent precast reinforced concrete constructions
took place, the latter have achieved the partial failure of the infills,
Fig. 81. (a) Localization of the plant VIBAC in the Bazzano municipality (AQ). (b) Sili before the earthquake.
RESEARCH - L’Aquila Earthquake
which have induced strong deformations of the shells of the sili. Such
type of damages are a clear effect of the earthquake vertical component,
whose importance they highlighted (Figure 82).
The Vibac sili have a metallic structure. Generally for their conception
sili have a very low structural weight, normally significantly lower than
the weight of the contained material. Such a characteristic implies a
very slender structure. It is evident that such structures are sensitive to
both local and global buckling phenomena. In fact the most common
failure mode is the instability of the wall panels due to the effects of the
axial force in compression. Such actions are due to the friction between
the silage material and the walls. The horizontal radial pressure, acting
on the cylinder surface from the silage material, has a stabilizing effect
against the buckling of the silos’ walls, giving rise to a tension stress
field of membrane type. The distribution and intensity of the internal
forces in every constituting part of the silos, the cylinder and the hopper, are strongly influenced by the material extraction behaviour, which
in turn depends on the shape of the silo.
The Vibac sili have an elongated shape typically used for the storage of
plastic material. Therefore the predominant extraction mode is of the so
called “mass” type, having the characteristic that the first material
coming out is the one inserted as first in the silos, all the material mass
is in movement at the leakage. In case of sili with a stocky shape, the
extraction behaviour of “funnel” type prevails, it has the characteristic
that a central tube forms in the material mass, which is sucked by the
hopper. Such a “tube” is fed by the silage material all along the height,
the part of material external to the tube rests during the leakage. In particular, in elongated sili, when completely full, along the height of the
cylinder, from the higher ring bands the radial pressure grows towards
the base up to assume a constant value, then at the section variation, it
reduces from the ring where the hopper is installed, and high stress
levels arise. Obviously in case of empty or partially full silos the behaviour is different, the stabilizing effect of the radial pressure is lost in
the empty part with a consequent abrupt variation of the critical stress.
Pressure variations inside the silos depend also by the leakage of the
material through the hopper, which causes a backwash effect and thus
depressions. In order to control and regulate such an effect, sili are provided with pressure valves. Given that it is plausible that on one side
the effect of the seismic vertical component provoked a sharp and
important increment of the actions in compression in the sili walls, causing buckling, the contemporary seismic action in all the components
accentuated the effect of possible asymmetrical distributions of pressure, due either to structural eccentricities, or to the silos filling method,
or to the anisotropy of the silage material, causing a reduction of the stabilizing effect of the radial pressures themselves. Furthermore, buckling could also occur due to constructional imperfections at the joints
between the coating ring bands of the silos, where joints in any case
represent a discontinuity in the flow of longitudinal stress in compression, with high concentrated stress. The above mentioned considerations fully justify the collapse behaviour observed during the L’Aquila
7.1 Road Network
ANAS S.p.A. is the agency that manages in the Abruzzo Region, as well
as in the rest of the national territory, the state road network. The residual functionality and safety investigation of the road network were the
first priorities identified by ANAS for the management of the first phase
of the emergency. Physical and human resources were deployed to
achieve the following goals: 1) rapid survey of the road network to ensure, at the largest possible extent, the regional mobility; 2) activation of
emergency contracting procedures (“somma urgenza” agreements) to
immediately begin, where possible, activities for the restoration of normal mobility conditions; 3) damage survey of the road-network components; 4) short term planning for the repair of damaged components.
At the same time, physical and human resources were deployed in support of the Civil Defence for a first partial debris removal and for the
excavations works necessary for the installation of relief campsites. It is
worth mentioning that, further to the local resources, additional ones
were used to manage the emergency. These resources were available
from few ANAS’ Regional compartments differently located on the
Fig. 83- Impact of the earthquake on the road network: (a) SS80 “Gran Sasso d’Italia” road affected by rock falls, but featuring
rock-proof tunnels. (b) Distribution of traffic management solutions (updated to 01/05/09) for the 61 road tracts affected
by the earthquake (red = road closed; dark green = passable with limitations; yellow = alternating one way; light green =
lane and velocity restictions).
national territory, with an average daily commitment of 80 men and 70
Rockfalls (Figure 83a) and landslides triggered by the earthquake and
aggravated by the heavy rain that hit the area in the days following the
event, were identified as the most problematic situations affecting the
network mobility. However, the rock falls and landslides occurred
mainly in mountainous areas around L’Aquila, while the main road
network in the city was not affected by the aforementioned phenomena.
In the urban area, mobility limitations were caused by debris following
damaged and/or unsafe residential and monumental buildings adjacent
to the roads.
Immediate activities for the restoration of normal mobility conditions
included: 1) removal of rocks and soil from the roads; 2) rock slope consolidations; 3) enhancement of soil slope stability. These activities were
conducted employing, where possible, internal resources or activating,
alternatively, emergency contracting procedures with external organisations. Securing of unsafe buildings adjacent to roads was carried out by
Temporary traffic management measures were extensively implemented
in order to minimize road closures; these measures included traffic flow
restrictions; alternating one-way; lane and velocity restrictions (Figure
The only significant damages occurred to the road network components
were the structural failure of the viaduct “Corfinio” on the national
roadway SS5 and the collapse of a bridge on the main road SP36
“Forconese”. No further significant damages were reported to the components of the road networks including the numerous tunnels present in
the Region that performed well.
The urgent need for a standardized and structured survey form to report
damages and disruptions in the road networks was highlighted while
performing safety investigation and damage survey operations. A rapid
survey form and an ad-hoc procedure were therefore identified and formalised while the survey work was in progress.
The timely information on the mobility conditions was a key component
of the effective emergency management. The Civil Defence issued daily
a report summarising road closures, mobility restrictions and repair
RESEARCH - L’Aquila Earthquake
works carried out in the road network. Using a Geographic Information
System, GIS, the technical compartment of the Direction of Command
and Control, Di.Coma.C represented this information in a cartographic
format. Road closures and other temporary traffic management measures were overlaid to aerial Figuregraphs, technical regional maps, etc.
providing maps that had a fundamental role in supporting many emergency management operations.
As for the public information, emergency bulletins were regularly
issued to update in real-time the end-users about the mobility situation
in the Abruzzo Region. Communications and timely news were, as well,
posted on the ANAS website.
Once the firth phase of the emergency was managed, efforts and resources were concentrated, on one hand, to handle the modified traffic conditions in L’Aquila city due to the closure of the main road that ran
through the city and, on the other hand, to respond to the new mobility
requirements created by the relief camps, and by the construction of the
provisional accommodation: Temporary Housing Modules M. A. P, and
C.A.S.E project.
7.2 Water distribution network
Gran Sasso Acqua G.S.A. SpA is the water provider for L’Aquila city
and for 37 municipalities in the earthquake area. The organisation
offers an integrated water service including potable water supply, sewerage and wastewater treatment.
The G.S.A. has 3 major supply systems (Chiarino, Gran Sasso, Water
Oria) in addition to some secondary ones. The water supplied is transported by a network consisting of approximately 900km of large diameter pipes and is stored in a huge number of tanks (about 200) that
require continuous functional and hygienic monitoring and maintenance. The water is distributed from the tanks to approximately 100000
customers through a 1100 km distribution network made of quite old
cast iron and steel pipes. The pressure inside the main pipeline network
is quite high, reaching 30-50 atm., as well as in the distribution
networks where it can reach 6-8 atm.
Thanks to a remote control service and guided valves connected, through
cables or wireless connection, to the main reservoirs and supply
systems, it is possible to check the water flow inside the pipeline
network and to manage partial or total opening/closing operations directly from the Gran Sasso Acqua headquarters. In particular, electromagnetic sensors, measuring input low pressure, and electromagnetic gauges (or “Clamp on”), measuring output differential pressures, are installed in the tanks. The remote control service allows furthermore the
assessment of the water level in the tanks.
The equipment connected to the remote control system revealed, on the
RESEARCH - L’Aquila Earthquake
morning of April 6, a significant and sudden change in the water flow
for a main pipeline in Paganica. The immediate closure of the relative
shutters for that pipe was operated directly from the GSA headquarters,
before the technician team reached the affected site. The cause of the
rupture was identified in the fault crossing the Paganica pipe. Because
of that, the steel joint of the pipeline (diameter = 600cm; pressure 2530atm) slip-off, causing a violent escape of water (Figure 84a).
A connection portion at the joint, however, was still grasped for a length
of 6cm. In order to quickly respond to the emergency, the repair was
limited to the welding of the pipes at the joint.
Exception made for the aforementioned joint slip-off, no significant
damage was observed to the main distribution and storage system.
Following the repair of the damaged joint it was, therefore, possible to
restart the provision of potable water for all municipalities administered
by the G.S.A. SpA since the evening of April 6. As a lot of ruptures were
expected in the minor water distribution system, in order to prevent
flooding and deterioration in the buildings already damaged, the decision was made, not to restore the water distribution in L’Aquila historical centre and in the most affected villages. For these areas, the restoration of the water provision was gradually operated starting from the
less affected zones and/or the zones with a strong need for reactivation;
priority was given to the strategic services, secondly to the commercial
and industrial activities, including the hotels to be reopened for the G8
meeting, and finally to the residential buildings classified safe, after the
specific AeDES survey. The partial restoration of the water distribution
was possible because of secondary networks and of a shutter system
that allowed the exclusion of areas where the water supply was not
urgently needed. A few days after the earthquake (19 April), due to a
further slip of the fault, the welded joint of Paganica pipe broke, requiring a further repair intervention.
The priorities identified in the second phase of the emergency management were, on one hand, the provision of the water service to the relief
campsites and, on the other hand, the management of all the activities
for restoring the water provision in L’Aquila City. To carry out the works
for the water network connection in the relief campsites, the technical
staff of the company (fully operative since the third day after the earthquake) was supported by the “Genio Civile” staff. On the other hand,
the works for repairing damages and restoring the functionality of the
water service in L’Aquila were operated, where possible, by the G.S.A.
SpA technicians, or activating emergency outsourcing procedures for
the most demanding operations. Relationships with external organizations have been unfortunately, nowadays, interrupted because of the
financial difficulties that the company is undertaking due to the lack of
Most commonly observed damages in the minor distribution system
Fig. 84- Impact of the earthquake on the water distribution network: (a) Joint slip-off in a main water network pipeline in
Paganica. (b) Repair on a cast iron pipe in a Paganica at the moment when some of the evacuated people were returning
were the slippage/breakage of the joints and the breaking of cast iron
pipes (Figure 84b). It is important to emphasize, however, that in large
part of the “red zones” (damaged zones with prohibited access) the
water network is still closed. Because of that, it has not yet been possible to completely estimate the extent and the spread of the damage suffered by the network1.
Finally, it is worth mentioning that the drinking water purity and quality has been officially tested and certified daily since the early days
after the seismic event. Because the G.S.A. official testing laboratory
was severely damaged after the earthquake, this service was guaranteed
via mobile laboratories of a neighboring water organization, C.A.M..
The third phase of the emergency management focused on the construction of the water distribution network and connections for the sites
identified for the construction of the provisional accommodation:
Temporary Housing Modules M. A. P, and C.A.S.E project. Both the
design and the new construction of the reservoirs and of the distribution
network for these areas were committed to external organizations and
contractors. The costs for both the design and the construction of the
new reservoirs and networks for the temporary accommodation were
covered by the Civil Defence. The G.S.A. SpA will continue to be in
charge of the management of the water provision for the temporary
accommodation areas.
7.3 Wastewater treatment plant
The technical visits at the wastewater treatment plants serving L’Aquila
(AQ), in the resorts of Ponte Rosarolo, Pile and Arischia, and at that
located in the City of Corfinio (AQ ) have shown that examined systems
have similar technical characteristics, as they have the same practical
functions. Each plant was equipped both with the structures necessary
for the treatment of wastewater (primary clarifier tank, aeration tank,
digestion tank, settling tank, thickener, sludge dewatering band press
The water consumption was reduced by 30% as a result of water shut off into the ‘red zones’.
Mobile water tankers were used to serve the relief camps in the first days after the quake.
and chlorinator system) and with those for management purposes
(buildings used as offices, rooms for technical equipment and laboratories).
The facility in Ponte Rosarolo is located near the historical center of
L’Aquila (42°20'18.10''N - 13°23'39.09''E). Structures were built the
’60s-’70s. The reinforced concrete digestion tank suffered partial collapse of a longitudinal wall (Figure 85a), several vertical cracks on a transversal wall and the separation of orthogonal walls at the edges (Figure
85b). The partial collapse of the wall also involved the steel pipe adducting wastewater that was connected to it. In buildings used as offices,
local technological and laboratory equipment (RC framed structure)
were also found cracks of both internal partitions and external walls.
However, there were no evidences of damage to structural elements: the
cracks detected on non-structural elements did not represent significant damages and did not prevent the use of building. The inspected
facilities were therefore useable at the time of inspection, except the
digestion tank that was useless. Due to this damage the tank has lost
water and the plant were partially closed by reducing the disposal capacity of about 60%. The remaining functionality was still sufficient to
face the demand, which was significantly reduced due to the large number of evacuated people (approximately 30,000), housed outside the
Fig. 85- Ponte Rosarolo Plant. Digestion Tank: (a) partial collapse of a longitudinal wall and of the pipe connected to it. (b)
Detail of the detachment of the orthogonal walls at the edges. (c) Displacement of the pump in the control room.
The structures of facility in Pile (42°21'3.25''N - 13°22'13.41''E),
which is situated between the town and the industrial area of L’Aquila
being the second plant serving the city, were realized in two different
periods (’80s and 2000) with RC walls and slabs. Structural damages
were not detected, only some damages to the partitions of local offices
occurred. With regard to the older settling tanks, characterized by a circular cross section, a deterioration of the curbing RC beam was detected due to significant corrosion of the steel reinforcements.
The inspected structures, therefore, were viable and fully functional
despite the damages (of non-seismic origin), due to degradation of materials descending from a insufficient maintenance of the settling tanks.
However, in the control room, a tube connected to the pump (not anchored) was damaged due to a displacement of 15 cm, Figure 85c. Finally
RESEARCH - L’Aquila Earthquake
it should be noted that this plant has been out of energy for three days
after the earthquake, so it worked through its own backup generator.
The plant located in Arischia (42°24'49.02''N - 13° 20'25.48''E) presents reinforced concrete structures with the exception of the circular
tanks for leaching, consisting of circular walls of artificial masonry
blocks connected with a RC curb at the top of the tank, and a gravity
retaining wall. The structures date back to the ’70s with the exception
of RC curb which was more recently constructed. Cracks on the walls
of a distribution trap and damages to the retaining stone wall, which led
to the partial obstruction of the hydraulic groove drain at the base of the
tank, were observed. With regard to the circular tanks, one of the two
rotating distributors was put out of service for damage to its support; the
cracks found on some blocks of the structure were dated before the
earthquake. Therefore, the inspected facilities were functional, although
the restoration of the full functionality of the hydraulic facility required
some minor rehabilitation and repair of the tank distributor. In any case,
the age of the plant suggests a constant monitoring even after the remedial action.
The treatment facility in Corfinio (AQ) situated not far from the center
of the same town (42°7'25.74''N - 13°50'31.78''E) is a RC construction
built in the ’90s. The central part of the longitudinal walls of the aeration tank, separated from lateral walls, shows a rotation very probably
occurred in large part before the seismic event, as witnessed by the
comparison of the positions of monitoring slides before and after the
earthquake; such slides were applied two years before the event: the
displacements due to the earthquake did not compromise the hydraulic
seal of the joint, nor the functionality of the structure.
A comprehensive analysis of the observed damages was carried out in
relation to the position of each facility with respect to the epicenter of
the earthquake of April 6th, 2009 (UTC 01.32 hours) and to the records
provided by the National Network accelerometric (RAN) available. It
can be observed that:
- Ponte Rosarolo facility is located near the epicenter and close to the
AQK accelerometric station, which recorded ground accelerations
equal to 3.7 m/s2 equal to about 50% of the maximum value recorded
for the same seismic event (station AGV - 6.6 m/s2); after the earthquake, the plant has shown damages to the tanks with rectangular walls
larger than those found in circular tanks of the Pile plant, despite the
geographical proximity. The structural behavior of the circular tanks
was essentially better than that of the rectangular ones, mainly because of the lack of structural details ensuring effective connection between the orthogonal walls;
- Arischia plant lies about 5 km from the L’Aquila accelerometric stations AQV, AQG and AQA, which recorded maximum ground acceleration values; even if distant from the epicenter (approximately 10 km),
RESEARCH - L’Aquila Earthquake
it has shown some structural damages;
- The Corfinio plant was not damaged because distant from the epicenter (approximately 50 km): the maximum acceleration recorded by the
accelerometers of Sulmona station (Sul) located near the plant, is indeed equal to 0.34 m/s2, approximately one-twentieth of the maximum
recorded at AQV Station of L’Aquila.
7.4 Gas distribution network
Enel Rete Gas S.p.A. is the gas provider for L’Aquila city and for other
5 municipalities in the earthquake affected area, namely Lucoli,
Tornimparte, Ocre, Rocca di Cambio, Rocca di Mezzo.
The gas is distributed via a 621 km pipeline network, 234 Km of that
with gas flowing at average pressure (2.5-3 bar) and the remaining 387
Km with gas flowing at low pressure (0.025-0.035 bar).
The medium pressure network is connected to the high pressure national
one (namely SNAM network) through 3 reduction cabins while, about 300
reduction groups allow for the transformation of the gas transport pressure (2.5-3 bar) into the gas distribution pressure (0.025-0.035 bar).
The gas network is mainly made of steel pipes, with an average internal
diameter of intenal =125cm (external diameter external =139.7cm) and
the joints are mainly welded.
The first priority identified for the management of the gas network, in
the first phase of the emergency immediately after the earthquake, was
the timely securing of the network in order to avoid explosions, gas
leaks and fires and to allow the emergency vehicles and the USAR
teams to act in the safest possible way.
To ensure this priority, the entire network managed by Enel Rete Gas
S.p.A. in the affected area was shut off via the closure of the 3 reduction cabins. Thanks to this decision, and to the rupture of a pipeline
near Onna (Figure 86a), it was possible to timely and significantly reduce the gas pressure and to avoid the occurrence of secondary effects.
The subsequent closure of the 300 reduction groups ensured the full
securing of the network in less than two hours after the earthquake. In
the days following the event, the gas valves external to each residential
building were as well closed. The pipeline damaged in Onna was replaced with a new one that was too rigidly connected to a reinforced-concrete support. It is worth highlighting that, as a result of the earthquake,
the Enel Rete Gas headquarters in L’Aquila resulted unusable. Because
of that the chief executive and the staff had to manage the emergency
without the support of their data, software and maps. Luckily, the national society Enel Rete Gas has, at a national level, an integrated information system, including a data base and a geographical information
system GIS. Making reference to the closest Enel Rete Gas headquarters in Teramo and Pescara, it was possible to reprint the maps and all
the documentation necessary to operate.
The second phase of the emergency response was focused on the activation of the physical and human resources in support to the Civil
Defence. The timely provision of gas to the strategic structures was the
first priority identified and was operated via mobile reduction cabins
and gas wagons. H24 shift were organized for the local technical and
administrative teams, as well as for the teams coming from other areas
of the national territory including the Enel Rete Gas national headquarters in Milan. In the first month after the earthquake, the daily
commitment of physical and human resources resulted on average
approximately equal to 70 men and 35 vehicles, including equipped
trucks, gas wagons and gas-leak detectors.
On the same time, activities for the reactivation of the gas provision were
started. The reactivation of the shut gas network required to operate gradually restoring, first of all, the gas flow into the medium pressure
network, secondly the gas flow in the low pressure network, up to each
external valve pertinent to each residential building previously closed.
Reactivation of the service was managed according to the following four
steps: 1) seal verification; 2) nitrogen check; 3) repair of damaged pipes
and/or valves; 4) reopening. In the seal verification phase, the detection
of broken pipes and/or the possible joint slip-off was made, acting in the
first instance, from node to node, and further segmenting the network
when necessary.
The material and equipment needed for the repair was immediately
available from the integrated logistics system which Enel Rete Gas
uses; actually, the material normally in storage in the Battipaglia interharbour to perform ordinary repairs and maintenance works, was simply diverted to L’Aquila. The adopted strategy ensured the remediation
and testing of more than 90% of the gas network in three month time
after the earthquake. The diagram in Figure 86b shows how, three
months after the quake, it was possible to restart the gas distribution for
all the end-users with a safe home, exception made for L’Aquila city.
It is worth mentioning that the reconnection of the individual user supplies required, on one hand, the definition of the priorities to be followed and, on the other hand, the definition of the testing procedures to
be carried out to certify the safety of the gas systems that were subjeca
ted to the action of the earthquake. As for the priorities, those identified
by the Civil Defence were followed; namely, the service was provided
first of all to the strategic buildings, secondly to the manufacturing and
industrial plants, and finally to the residential buildings identified as
safe after the AeDES ispection. As for the testing procedures, in accordance with the procedures used by Enel Rete Gas for routine checks,
an ad hoc protocol was defined in collaboration with the Civil Defence
and the Firefighter Department. It was decided to reconnect each single user following the fulfillment of four conditions: 1) safe dwelling
(classified as A following the AeDES survey); 2) leak-tightness
checking; 3) operative test of the equipment; 4) smoke test. It is worth
mentioning that the Civil Defence fully covered the cost of the whole
procedure to reconnect the individual users to the gas service and that
a dedicated phone line (Line Amica Abruzzo) was specifically set up to
facilitate and support the end-users in this operation.
As a final note it is worth remembering that no damages were detected
to the gas storage facilities.
7.5 Electric power distribution network and telecommunications
It was reported that two substations serving the greater L’Aquila had
damaged connections between a rigid bus and insulator, Figure 87a.
That was due to shifting of the un-anchored transformers during the
earthquake. Also due to sloshing of the cooling oil within the transformer, cooling oil pressure increased, and actuated the safety shut off feature to avoid costly damage. One of the transformers moved about 14
cm. In the distribution system, 30 posts were damaged causing severed
links that resulted in service disruption. More than 180 pedestal type
connection boxes were dislocated and severed cable connections at the
termination lugs that resulted in localized power failure (Figure 87b).
The Electric Power Control Center at L’Aquila sustained severe damage, both building and equipment, and it had to be moved to a temporary
building in the yard of the building premise. It took three days to complete the move, while the essential part of the system was functional by
9 AM the day after the earthquake (Figure 86). Transformers in substations were not anchored. We noted that steel angles were welded on the
Fig. 86- Impact of the earthquake on the gas distribution network: (a) Onna (AQ),
damaged pipeline. (b) End-user gas connections activeted on June 8, 2009
(Green = end-users that can be potentially reconnected; Bleu = end-user
reconnected with respect to total that can be potentially reconnected).
RESEARCH - L’Aquila Earthquake
Fig. 87- (a) Damage to rigid connection of a transformer. (b) Typical damage to pedestal box.
Fig. 89- One of the relief campite in L’Aquila set up by the Civil Defence.
Fig. 88- (a) Steel anchors installed after the earthquake to avoid sliding of transformers. (b) Unanchored batteries’racks in
tracks that the transformers were supported to stop sliding, Figure 88a.
This was done after the earthquake. However the steel angles seemed
to be under sized. In the control house of substations, the batteries were
not anchored or tied to the racks, Figure 88b. There was no batteries
damage reported at these substations. Some locations were without
power for three days, e.g. wastewater treatment plant.
Telecommunication service performed reasonably well. It went off air
for a couple of hours right after the earthquake. Cellular phones seemed
to be the main means of telecommunication in this small community.
Although there was no reported damage to the physical equipment and
equipment building, we saw a number of temporary cellular sites
deployed within the earthquake impacted areas. The increase of cell
sites might have reduced the circuit overload that commonly occurs
after an earthquake. Both Fire Fighters and Police used their own radio
system as the primary communication tool. Cellular phones were also
used to compliment the radio system. With a good backup power generation plant, their communication was not interrupted. The Fire department had three repeater stations, which were not damaged. A number
of landlines were damaged or severed, as repairs were evident during
C.F. Carocci, S. Lagomarsino, Masonry Buildings in the historic centers of the
L’Aquila area, Progettazione Sismica, 2010. IUSS Press, Pavia.
E. Cosenza, G. Manfredi, G.M. Verderame, Reinforced concrete buildings,
Progettazione Sismica, 2010. IUSS Press, Pavia.
M. Menegotto, Observations on precast concrete structures of industrial buildings and warehouses.
M. Di Ludovico, G. Di Pasquale, M. Dolce, G. Manfredi, C. Moroni, A. Prota,
Behavior of scholastic buildings after L’Aquila earthquake, Progettazione
RESEARCH - L’Aquila Earthquake
our investigation. Since tenants were not allowed back to their houses
or apartments, most landlines were not used. Hence the demand on this
circuit became much lighter.
7.6 Temporary housing
The Italian government organizations and NGOs (Non-Government
Organization) were to be commended on a great effort providing the victims with relief services and care. The military and fire brigade set up
service camps to provided needed services to the victims. Some of the
relief campsites provided the victims with Internet services in addition
to daily necessities such as medication, food, and water. In general the
victims were very satisfied with the relief service. Many residents were
afraid to get back to their houses even when their houses (marked as
class A or B) were not condemned, due to their fear of future earthquakes and the potential for damage to their homes. Temporary housing
is scheduled to be completed by September 2009 (before winter arrives)
for the victims, Figure 89. These houses will be on a base isolation
system to protect residents from future earthquakes. There were more
than 30,000 victims settling in more than 160 campsites.
Sismica, 2010. IUSS Press, Pavia.
C. Casarotti, A. Pavese, S. Peloso, Seismic Response of the San Salvatore
Hospital of Coppito (L’Aquila) during the 6th April 2009 earthquake,
Progettazione Sismica, 2010. IUSS Press, Pavia.
B. Faggiano, I. Iervolino, G. Magliulo, G. Manfredi, I. Vanzi, Post-event analysis of industrial structures behavior during L’Aquila earthquake, Progettazione
Sismica, 2010. IUSS Press, Pavia.
M. Dolce, S. Giovinazzi, I. Iervolino, E. Nigro, A. Tang, Emergency Management for lifelines and rapid response after L’Aquila earthquake, Progettazione
Sismica, 2010. IUSS Press, Pavia.
Reconstruction between temporary and
definitive: the CASE project
should actually be considered permanent, since they had a lifetime longer than 10 years (ignoring the fact that apparently between 10 and 50
years works can neither be called provisional, nor permanent).
If then the provisional does not exist from a durational point of view, it
would be useful to wonder whether it makes sense that it would exist
looking at energy consumption, sustainable environment or pollution. It
would also be useful to wonder whether buildings could be constructed
with environmental characteristics and safety level similar to that required for permanent ones on a temporary basis and with cost per unit
similar to provisional ones. If this should be the case, it would be logical to propose to build provisional houses with characteristics of the
permanent ones.
These ideas and others were discussed in the days directly following the
Aquila earthquake with Guido Bertolaso for the political, administrative and economical aspects, with Mauro Dolce, Edoardo Cosenza and
Gaetano Manfredi for the technical and scientific aspects.
A first complete conceptual proposal, with 3D-rendering and preliminary calculations was submitted on April 16th, together with several
comments. It was hypothesized to deliver the buildings for 3,000 inhabitants within 5 months, guaranteeing seismic safety by means of an
hat is the time difference that distinguishes a temporary or provisionally home from a permanent or final? It is not easy to
respond to this question, if you consider the seemingly enduring eternity of what in Italy is built with the objective to last for months, or for
a maximum of few years.
With reference to Italy, it is enough to consider what happened after the
earthquakes of Belice and Irpinia (or even in Friuli), there is therefore
no need to further elaborate the concept.
On the other hand, we could refer to the technical code of 2008 [1], in
which the nominal lifetime of a structure is defined as the number of
years in which the structure – normally maintained – can be used for the
purpose it was built for, it is indicated in a table and it needs to be specified in the design documents.
It is interesting to note that the code only indicates a maximum for provisional works (10 years) and two minima for ordinary and important
works (50 and 100 years respectively). If one sticks to these data, it
should be concluded that all the provisional works that were constructed in the aftermath of the earthquakes that took place after WWII
Fig. 1- One of the first sketches of the project illustrating the logic of the buildings constructed on isolated plates.
Fig. 2- A plan sketch made by architects Ragazzi e Hoffer as to illustrate the logic of the infrastructure in a court
open to pedestrians.
Fig. 3- One of the many 3D renderings used to illustrate design hypotheses.
RESEARCH - L’Aquila Earthquake
Gian Michele Calvi1 and Vincenzo Spaziante2
Eucentre Foundation - Centro Europeo di Formazione e Ricerca in Ingegneria Sismica, Pavia.
Department of Protezione Civile, Rome. www.protezionecivile.it
isolation system at the level of an urban block, and proposing elevate
standards of living, technology and environment protection. The pursuit
of these objectives, apparently impossible, was based on the construction of large isolated plates and the subsequent assembling of pre-fabricated three-storey living units. The need for the project to be as much
as possible independent from local soil conditions and from the unknown construction technology (many different ones would have been
necessary to meet the deadlines) became immediately clear. To this end
it was stated the need of urgently identifying the possible building technologies compatible with the timing programme and the technical constraints, of selecting technical and commercial partners and of explo-
ring the production capacity of the market. The time programme was
defining in four weeks the date to open the construction sites, i.e. to
start construction by mid-May, to deliver houses to 3,000 inhabitants by
The economic analyses indicated an estimated cost of 120 million euro,
VAT excluded, for 3,000 inhabitants, with a 20% uncertainty rate and
without considering furniture, purchase of the terrain and photovoltaic
In preliminary calculations it was assumed to use friction pendulum
devices [2-8], with a radius of curvature of 4m, a vibration period of 4s,
a displacement capacity of about 300 mm, a friction coefficient between
Fig. 4- Images used in the preliminary phase to illustrate possible technologies for the assemblage of the buildings.
3 and 5% and an equivalent viscous damping between 20 and 25%.
The alternative of using rubber isolators was also taken into consideration, but appeared in this specific case to be less competitive, considering the relatively low axial forces and the large horizontal displacement
In the days immediately following, several aspects that would have permanently defined the project were discussed and clarified:
- The reduction of each one of the isolated plates to about 20 by 60 m,
suitable to sustain three-storey buildings with each floor surface of
about 600 m2, with a capacity of about 80 inhabitants in 25 to 30 apartments. Plates of this dimension should allow an adequate flexibility in
relation to the plane altitude conditions of the areas to use (at that
RESEARCH - L’Aquila Earthquake
moment unknown) and the construction technologies, also unknown;
- The definition of 150 as the approximate number of plates to construct
and therefore of about 12,000 inhabitants to settle in;
- The division of the intervention in numerous small villages, consisting
of 4 to 20 plates and hence a number of inhabitants between 300 and
- The definition of a serial timesheet in which a group of 30 plates
should be finished about 15 days after the previous group, which
implied a forecast of delivery of the apartments in 5 tranches for 2,400
inhabitants a time, with deadlines spread out between 30 September
and 30 November;
- The decision to manage the entire project directly, without intervention of a general contractor, setting up a non-profit technical structure
that responds directly to the Civil Protection Department (DPC). It was
thought that this way it would be possible to save substantial economical resources, mainly on general additional costs and to have a more
accurate control on deadlines and quality of the project.
The definition of an operational, management and outsourcing structu-
re, of personnel roles, activities and their interaction, time programme
and milestones required several days of intensive work and was completed and formalised by May 8. The way the project is managed is very
innovative with respect to the schemes that are normally adopted, and
not only in Italy. In fact a single–purpose consortium was created
(named ForCASE), formed by Eucentre (a non-profit foundation, centre
of competence for seismic risk of the department of civil protection,
founded by four public institutions and with a nature of ‘public company’ in Europe) and two construction companies, (ICOP and Damiani).
The two companies agreed to operate in this context as non-profit entities and not to participate in any other reconstruction activity in
Abruzzo. Their role would have been that of a technical office, and therefore to facilitate the consortium to act on behalf of the CPD as a general contractor, with the capacity to manage directly supplies purchasing,
to coordinate activities on the construction site, to arrange and verify all
accounting matters.
Obviously, the consortium had as well the main task of carrying out all
designing and construction management activities, under the responsibility and coordination of the authors of this article. Coherently with
what has been briefly described, the operational organogram demonstrates five main areas of actions: two being related to design activities,
Fig. 5- The personnel and work organization plan set up in the preliminary phase.
RESEARCH - L’Aquila Earthquake
pletion of the project;
- Finalisation and publication of call for bids;
- Stipulation of contracts;
- External checking and control;
- Relations with institutions and obtainment of permissions;
- Identification of the intervention construction sites, expropriations of
lands and related activities.
Fig. 6- A simplified version of the extremely detailed and complex time schedule that allowed daily overviews on each
aspect of the project and the construction.
two to management and accountancy activities and one to project coordination.
In order to obtain maximum efficiency, in terms of time and costs, and
to ensure quality control, three different operational modalities for contracting and execution of the work were identified:
- For the activities of preparation of the construction site and infrastructure works, it was decided to mainly use local contraction companies;
- For the foundation and isolation systems, it was opted to act directly
as a general contractor acquiring materials and supplies, such as concrete, welded wire meshes, steel columns, isolation devices, formwork
positioning, concrete casting, etc.;
- For the construction of the housing structures it was decided to launch
a call for bids that included final design and global construction,
allowing the use of any building technologies compatible with the needs
and available time, and selecting the proposals with the highest quality
and the lowest cost.
The economic quantification of the costs for the management of all activities was estimated based on the pure cost of the staff assigned to this
temporary job (in months), on a monthly cost, in general between 3,000
and 12,000 euro (these are costs for the consortium, not net salaries),
and on a sum to cover cost of accommodation and travel, that could in
any case never exceed 3,000 euro per person per month.
As all the activities would be executed within the non-profit framework
that characterizes the Eucentre Foundation and the ForCASE consortium, the estimations were considered as a maximum not to be exceeded, while the real costs would be subjected to accountability checks.
The Department of Civil Protection would directly execute, in cooperation with the consortium, all the activities related to:
- Definition of Civil protection ordinances, possibly needed for the comRESEARCH - L’Aquila Earthquake
The architectural project of a building unit, as briefly presented, favoured the development of different types of apartments, as a function of
family compositions, which resulted in 109 different shapes after the
selection of 16 contracting bids, as discussed later.
Regarding the choices on infrastructure, it needs to be highlighted that
a first guiding concept was that of placing the settlements in the neighbourhood of existing villages that had suffered severe damages because of the earthquake, to be able to relocate the people within their own
territory, to preserve the close ties that people have with land and neighbours.
This general principle was confronted with technical difficulties deriving from non-ideal geomorphic, hydrological and geotechnical circumstances of the areas, to finalise the best possible selection of the
areas of intervention.
Once the settlements had been defined and sized as a function of quantitative needs and land capacity, considering the dimensional and
morphological characteristics of the location, the problem of existing
infrastructures (roads, pipelines, sewing system, etc.) and of their
improvement and integration had to be faced.
Finally, the population indices could be defined, starting from figures
between 100 and 150 inhabitants per hectare, for location in more rural
or more densely populated areas. Such figures imply a rather sparse settlement typology, marked by large green areas.
A final infrastructure index had been identified by assigning 30% of the
land surface to services and facilities, such as leisure, sport, shopping
centres or education and religious structures.
Based on these premises the final urban design of the areas was completed, obviously combining the building units previously described
(essentially consisting of three inhabited floors above a covered parking), also considering exposure to sunlight, valley and mountain views,
steepness of terrain.
Driveways and walking paths were kept separated to the maximum possible extent, generally locating vehicles roads on the outer skirts of each
area, with access limited to parking lots and ground floors of the buildings, also used as parking. The walking paths were designed elimina127
Fig. 7- An example of the plans that were
designed for the bids of housing construction, with an underground parking between the two plates.
RESEARCH - L’Aquila Earthquake
Fig. 8- Location of the construction sites, all within the municipality of L’Aquila.
Fig. 9- Examples of the urban plans for some site.
Coppito 2
Fig. 9- Examples of the urban plans for some sites.
RESEARCH - L’Aquila Earthquake
ting all architectural barriers, connecting green areas and inhabited
levels to road systems and parking lots with external elevators when
The final character of the new settlements tried to combine in an optimal way people needs, environmental and landscape requirements, use
of existing infrastructures and construction of new ones, in an integrated vision.
Later, another problem had to be faced, i.e. how to combine each one of
the specific building units (at this stage 150, in 20 different locations)
to each one of the 16 different typologies proposed by the companies
who won the call for bids. Choices had to be made in relation to construction technology and material, external aspect, number of buildings
awarded to each company, construction plan and schedule proposed by
each company.
Finally, considering the high environmental value of the landscape, the
design and realisation of the green areas was again the subject of a
public, international call for bids, where again cost, time and quality of
the proposal were considered to select the winning bids.
Preliminary considerations
The structural design of the buildings constitutes the fundamental element that allowed the development of the entire project and is extremely simple in its basic logic: two reinforced concrete plates, separated
by columns and isolators, the lower one being in contact with soil and
the upper one with the building. The plates were designed without
knowing the local soil properties, nor the weight and plan distribution
and structure of the buildings. Therefore for both aspects conservative
assumptions were used, to be verified later. In a few cases, some potentially selected construction location had to be discarded because the
soil properties appeared to be unsuitable.
It should be noted that the two plates are characterized by similar flexural actions induced by gravity, if it is assumed a uniform distribution of
the building load and of the soil reaction. Preliminary evaluations, based
on a column span of 6 m in both directions (convenient for parking arrangements), lead to a required thickness of both plates of 500 mm.
The weight of each building, with three floors of about 600 m2 each, was
estimated in a maximum of 21 MN, with a consequent total maximum
weight of slab, dead loads and building details of between 30 and 40
MN (or an average weight per column of about 1 MN).
The first vibration period of the building can be estimated between
Ts = 0.25 and Ts = 0.45 s, using the equation:
Ts CsH0,75
in which H is the height of the building and Cs is 0.05 for wall structures, 0.075 for reinforced concrete frames, 0.085 for steel frames. It is
however well known that equations of this sort tend to underestimate the
real vibration period resulted from a secant stiffness to yield, that for the
examined buildings could arrive at values between 0.8 and 0.9 s [9, 10].
Based on these considerations, the design period of vibration of the isolation system was selected in the range of 4 s.
It was also preliminarily observed that even an extreme temperature
variation of ±30 °C, leads to variations in length of about 8,5 mm on
each side of the axe of symmetry, that would not induce excessive horizontal loads into the columns.
Seismic action
Seismic action and in particular spectral demands in acceleration and
displacement are discussed in detail elsewhere in this volume [11].
Here it is however important to note that the fundamental parameter to
be assessed for a proper design of the isolation system is the maximum
displacement demand at a period of about 4 s. The spectra derived from
the registrations of April 6th show generally displacement demand of
less than 120 mm, with one exception, the AQK registration, in which
spectra values are close to 250 mm. The code spectra for events with
return periods of 1000 years, to be used for the design of the isolation
system, have values of about 300 mm for soil type B and 400 mm for
soil type E. These values can be significantly reduced in presence of
energy dissipation, as a function of an appropriate equivalent damping,
according to the factor:
where is the equivalent viscous damping value, that could be in the
order of 10-15% for rubber bearings and of 20% for friction sliders. The
values obtained from the reduction coefficients are between 0.6 and
0.7, with consequent estimations of displacement demands of about 250
mm for soil type E.
For the non-linear analyses the code spectrum for vertical actions has
also been considered, while for the building phases it was defined as a
‘construction event’ consistent with what indicated in addendum A of
Eurocode 8, part 2 [13]. Such an event appeared to be consistent with
registrations corresponding to a magnitude of 4.0, and was thus considered reasonable. While the demand in terms of acceleration was significant (in the order of ag = 0.10 on stiff soil), the displacement demand
was negligible.
Eight sets of spectrum compatible accelerograms have been used for
non-linear analyses, derived from registrations made in L’Aquila (3
records), and during the events of Imperial Valley in 1979, Loma Prieta
RESEARCH - L’Aquila Earthquake
Fig. 10- Acceleration and displacement spectra of an event with a 1000 year return-period in L’Aquila, according to the Italian code [1], soil category B and E, damping 5%.
in 1989, Northridge in 1994, Kobe in 1995 and Taiwan in 1999 (one for
each of these events).
Isolation system
The design and the verification of the isolation system was carried out
considering the possibility of adopting two different configurations, characterised by different devices, one based on the use of 12 elastomeric
isolators, together with 28 multi-directional sliding pot-bearings and
the other on the use of 40 isolators sliding on spherical surfaces, universally known as friction pendulum [FPS, 2].
Both choices are compatible with the project requirements, in different
ways. Actually, the smaller dissipation capacity of the system with elastomeric isolators (estimated to be equivalent to 12% damping) with
respect to the one with FPS isolators (estimated damping 20%,) requires a larger displacement capacity; in the order of 300 to 360 mm for
the elastomeric isolators, versus 260 mm for the FPS, depending on the
soil properties.
Obviously other combinations may be possible, also related to the
various displacement demands for isolators placed in different positions
(because of the eccentricity of the loaded mass, even only accidental,
the demand at the perimeter is larger than that closer to the slab central area). It was therefore allowed to bidders to propose different solutions, provided that they were respectful of design performances and
input. The result of the call for bids, in which FPS systems were preferred, should not be considered as a general demonstration of superiority with respect to elastomeric devices, but rather as a consequence of
the specific conditions of this project, characterised by relatively large
horizontal displacement demands, low vertical forces on the devices
and relatively low horizontal stiffness (as discussed, vibration periods of
the order of 4 seconds were assumed). This was the reason why elastomeric isolators had to be coupled with pot bearings: the use of rubber
bearings alone would have resulted in stiffness values incompatible
with the requirements of the project.
In the case of the FPS devices, the force corresponding to a displaced
position is defined by the following equation:
F = Mg + Mg d
( )
Fig. 11- Comparison of several spectra recorded on April 6th on soil type B, code spectra for an event with a 1000 year
return-period according code [1] and results of a recent research project (DPC-INGV-S5 [2]).
RESEARCH - L’Aquila Earthquake
In which Mg is the axial action (M is the mass and g the acceleration of
gravity), R = 4 m the radius of the spherical surface, = 3% is the friction coefficient and d the displacement of the isolator.
The least favourable conditions for the verification of displacement
capacity of the isolation system versus the corresponding demand are
likely to be those of a rigid and heavy superstructure, i.e. those of a
large participating mass and deformations concentrated in the isolation
system. With a configuration of this sort, the system global characteristics (40 pieces) resulted to be as follows.
Fig. 12- Force – displacement response of a system of 40 isolators and heavy superstructure.
Fig. 13- Force – displacement response of the system considering axial force variation due to vertical acceleration and global
interaction response [2, 4].
Effective stiffness, secant to the design displacement:
Verification of slabs and columns
The foundation and isolation plates have been subjected to numerous
finite element analyses, that allowed to calculate the maximum bending
and shear demand levels for several load combinations, to design the
reinforcement, generally made by welded wire meshes to favour a fast
positioning, and to verify the resulting action combinations with appropriate strength domains.
Local verifications for loads concentration at the column ends were also
performed on both slabs, considering the consequences of the substitution of a bearing as well. This operation was needed in hundreds of case
during construction, when the isolators were not yet available at the
time of casting the upper slabs.
The columns have been designed and verified considering either the
case of reinforced concrete and of steel, again to allow the use of various
Keff = 14615 kN/m
Corresponding period of vibration of the isolation system (note that in
general heavier structures are also stiffer, therefore characterised by
lower vibration periods):
T = 2
M =3.29s
Corresponding equivalent damping:
FPS = 2Mg = 0.201=20.1%
Fig. 14- Examples of displacement histories for an elastrometic isolator (left) and for a FPS isolator (right), subjected to events with a 1000 year return period derived from 3 registrations in L’Aquila, compared with capacity circles of 360 mm (left) and
260 mm (right).
RESEARCH - L’Aquila Earthquake
Fig. 15- Examples of reinforcement distribution in a section of the isolated plate.
technologies and thus reducing the operation time. For the same reason
steel columns were in general preferred, even if more expensive, using
concrete only when no steel elements were ready to be mounted.
Prescription for building design
As already mentioned, the final design of the home buildings was left to
the bidders, to allow the use of any building technology. However, the
specifications to which the projects would have anyway to conform needed to be defined as to assure an appropriate safety level to the global
structural system.
The seismic demand was defined in terms of design acceleration of the
building masses, calculated with reference to the maximum value of the
ratio between base shear and weight of the building, obtained in the
worst loading conditions, corresponding to those of a stiff building (T =
0.19 sec) with the lowest mass (1500 t).
For analyses performed with accelerograms compatible with the design
spectrum at a collapse limit state (SLC, return period 1000 years), the
base shear always resulted less than 0.11 times the weight of the building. It was therefore prescribed to assume a design acceleration equal
to 0.1 g to verify the buildings at a life safety limit state (SLV, return
period 500 years).
For the same conditions, the average inter storey drift resulted on the
range of 0.1%, a value that certainly allows a full use of the buildings
even after a high intensity event.
Together with these extremely simple design data, a series of prescripRESEARCH - L’Aquila Earthquake
tions on the characteristics of the buildings needed to be defined, in
order to keep them within the parameters assumed for the analyses and
therefore avoiding unexpected responses and jeopardizing the verification of plates, foundations and isolation system. A summary of these
prescriptions follows:
1. The load resulting from the building structures shall not induce in
any element of the slab – foundation system local actions larger than
those resulting from a uniformly distributed live load equal to 50 kN/m2
(i.e. excluding the slab self weight).
2. The load distribution on the plates shall exclude concentrations
potentially resulting in local collapses.
3. The maximum vertical action on a single bearing should be less than
2800 kN, either for the seismic load combination and for the gravity
combination at the ultimate limit state, including the weight of the
4. Bearings shall not be subjected to tensile forces in any load case.
5. The main period of vibration of the building (considered fixed at the
base) shall not exceed 0.5 seconds.
6. The eccentricity between centre of mass of the building and centre of
mass of the plate shall be less than 5% of the total length of the plate
(57 m) in the longitudinal direction, and less than 10% of the length of
the plate (21 m) in the transversal direction.
7. The maximum seismic mass of the building alone (i.e. without considering the weight and loads of the slab), calculated including self
weight, dead load and the fraction of live load to be considered for seis133
Fig. 16- Example of bending stresses in the foundation plate, for gravity
loads (1st row, moments around the two axes of symmetry in kNm/m)
and for seismic loads (2nd row: maximum values, and 3rd row: minimum
Fig. 17- Example of local reinforcement of the foundation plate at column bases.
Fig. 18- Example of bending moment – axial action strength domain, for a section of plate (at
columns centres).
Fig. 19- Example of bending action on the isolated plate during bearing substitution at different
RESEARCH - L’Aquila Earthquake
Fig. 20- Reinforcement in a foundation plate with concrete columns and steel columns with isolators on a casted plate.
mic verification shall be less than 2100 t.
8. The buildings shall be designed in accordance with the technical
code of 14/01/2008. It is accepted to represent the horizontal load equivalent to the seismic action by means of a static force vector, to be
applied to the building floors, according to equations given in the code,
assuming a design acceleration Sd (T1) of 0.1 g.
As previously discussed, for the production of the plates, the ForCASE
consortium has directly taken the role of general contractor, with calls
for bids for excavations, supply of concrete (initially about 200,000 m3,
with peaks in delivery of more than 5,000 m3 per day, self compacting
and aerated), supply of welded wire meshes (initially about 260,000 kN,
in general with diameter 14 mm at 100 mm), supply of steel columns
(initially 180,000 ton, diameter 800 mm), supply of isolators (initially
6,000 pieces, including assistance to positioning) and supply of casting
forms (initially for about 336,000 m2) and on-site assistance for reinforcement positioning and pouring of concrete. All quantities were later
significantly increased, since the number of buildings passed from 150
to 184. The prices per unit obtained through bidding have been the following:
Contractors for the production of the plates with initial price and offers
RESEARCH - L’Aquila Earthquake
Fig. 21- Rendering and floor plans of some buildings, proposed by the bidders.
RESEARCH - L’Aquila Earthquake
Contractors, structure material, number of buildings offered and price per building
• Self-compacting concrete
• Welded wire mesh
• Steel columns
• Isolators
• Forms and on-site assistance
A public call for bids was launched for the construction of the buildings; including final design. The 150 buildings to be built were grouped in 30 lots, each one of 5 buildings, allowing a bidder to present a
proposal for a maximum of 10 lots.
Depending on the final ranking of the offers, it might have been possible to have from a minimum of 3 contractors (in case the first 3 would
each propose 10 lots) to a maximum number of 30 contractors (in case
each one would have proposed 1 lot). The basic price for any lot of 5
buildings (about 160 covered parking spots, 3,000 m2 of outside pavement and 9,000 m2 internal living area) was fixed at 11 million euro.
The evaluation of the proposals was essentially based on the proposed
improvement of the minimum performance characteristics foreseen by
RESEARCH - L’Aquila Earthquake
the existing norms (that already represent a high standard). The maximum time allowed for completing each building from the availability
of the upper plate was fixed at 80 days, a proposed reduction was also
considered in the evaluation, together with a reduction of the proposed
Following the presentation of the 58 proposals and an accurate review,
16 contractors were selected, with a total average amount per lot of
about 10,500,000 euro, which means an offered price reduction of
about 5%.
On a total of 150 buildings, timber structures were proposed for 75
(50%), concrete structures for 45 (30%) and steel structures for 30
To complete the project it was necessary to prepare and launch other
5 groups of bids, in order to satisfy various needs:
- The upgrading and integration of the external infrastructure
(networks of any type) with the difficult problem of interacting with the
Fig. 22- Same complete buildings.
construction sites. Twenty bids were released, one for each area of
intervention, inviting companies located in Abruzzo and preferably in
the province of L’Aquila. Five companies randomly sorted out were
invited to bid for each site.
- The furniture and supplies necessary to immediately use the apartments. In this case a public competition was set up, on four lots of
about 1,000 apartments each. The foreseen time to assemble the furniture on site was 6 days from the moment an apartment would be finished. 18 companies presented an offer, with the following four resulting winners: Deltongo Industrie spa, Mobilificio Florida srl, RTI
Europea spa – P.M. International Furnishings srl – Martex spa and
Estel Office spa. The average price reduction offered was about 34%,
which corresponds to an average cost for the interior furnishing of an
apartment of 9,500 euro. It has to be underlined that the specifics of
the bid requested the highest possible standards also for the electrical
and mechanical equipment included in the offer, such as dish washer,
washing machine, tv set, etc.
- 309 elevators to connect the various floors of the buildings and 129
elevators to connect the buildings to the parking ground floor. This
need derived from the specific choice of completely eliminating all
potential architectonical barriers case, in excess of what compulsory
for legal requirement. A call for bids was released for three lots of 146
elevators each; 12 companies participated in the competition.
Marrocco elevators srl, ATI S.A.S. srl – Grivan Group srl, Schindler
spa resulted winners with an average reduction of price of about 16%.
- The opportunity of producing electric energy on site, collocating photovoltaic panels on the roofs. An estimate of about 45,000 m2 of roof
surface was considered adequately exposed to sunshine and consequently another bid was released for the design, construction, mana138
gement and maintenance of a photovoltaic system, capable of producing about 4,500 kW. The call for bid assumed that there should have
been no cost to the administration, and was based on technical merit
and on one fundamental economical parameter, i.e. a yearly fee to be
paid, as a percentage of the public incentives provided to favour the
use of alternative, renewable energy sources. The winner and therefore contractor was Enerpoint spa, Ener Point Energy Srl and Troiani &
Ciarocchi Srl., who offered to refund 9,01% of the incentives.
- Finally, two last calls for bids were launched to complete the green
areas, simply grouping the eastern and western construction sites. The
offers should obviously include land preparation, grass, bushes and
trees, walking and cycling paths, but also irrigation and drainage
systems, external furniture, sport and leisure fields. 19 companies participated in the bid and contractors were selected on cost and on evaluation of landscape beauty, environmental sustainability and maintenance and management characteristics of the offers. The selected contractors were 3A Progetti S.p.a, which lowered the estimated price of
39% and the Sestante Consortium, that managed to offer a 35,16%
The extremely limited time available for the completion of the project
required an extreme level of control and programming, with a continuous flow of information between engineering work management and
construction companies and daily reports and checks.
A coordination and management system was therefore set up, focusing
on the definition of priorities and of the main activities consequently
RESEARCH - L’Aquila Earthquake
Fig. 23- Work in progress and completed
RESEARCH - L’Aquila Earthquake
required and on the identification of potentially critical elements and
work phases that could threaten the fulfilment of the programmed time
As previously discussed, the work activities and their management and
control were organised in five sectors, corresponding to production of
the plates, buildings (including interior design and furniture), mechanical and electrical installations, infrastructures roads and green areas.
For each technical sector a technical coordination structure was defined with a responsible for programming, coordination with the construction companies and management of works. The general timeline of
the works was accordingly subdivided into the same five sectors. Daily
updates on the work progress and comparison with the time planning
guaranteed that each sector was closely monitored in terms of work progress, as well providing all technical personnel with an overview of the
general picture, fundamental to manage the coordination between different sectors. The graphical visualization of the daily progress of the
works resulted to be particularly useful for a rapid interpretation of the
complexity of the data that needed to be managed.
An idea of the large quantities of materials and labour that needed to be
managed, can be obtained by considering the example of the foundation
and isolation system, that on a daily basis needed an average concrete
Fig. 24- Example of a global daily overview form.
Fig. 25- Example of a production summary overview daily form (25 October, general overview on the left, production and delivery of plates on the right).
RESEARCH - L’Aquila Earthquake
Fig. 26- Example of a daily report on the general development of the project
works (19 October).
Fig. 27- Example of a daily report on the general developments of the works in
a specific area (19 October).
RESEARCH - L’Aquila Earthquake
supply of the order of 5,000 m3, welded meshes and reinforcing bars for
about 10,000 kN, about 200 steel columns (with a diameter of 800 mm),
to be provided in general in 20 different construction sites.
The efficiency of the team that was set up to program and coordinate
the activities, allowed such a proper and precise forecasting of the
work progress that all construction sites proceeded always on time and
actually all works were completed ahead of time, despite of the difficulties inherent in the number of workers (more than 8,000 in some
phases) and in the complex interaction between different work activities. An example of the time schedule programmed for the construction
of the two slabs systems, with foundation and isolation, can be summarised in Fig. 23.
Finally, the great efficiency of the team in charge of controlling all
aspects of safety in the work process should be noted. The extremely
detailed and continuous checks allowed the completion of hundreds of
millions of euro worth of work in just a few months without any notable accident and with the appreciation of all external controlling institutions, from those aimed to assure workers health to the unions.
The total cost of the project is split in the table below, considering the
different category of work and giving the average cost per building, per
apartment and per square metre of living space.
The total cost of 655 million euro refers to a total of 164 buildings,
while 150 were foreseen at the starting of the project and 184 were
actually built at the end. Applying a criterion of linear proportion it can
be inferred that the original 150 buildings would have cost 599 million
euro, which is in line with the 700 million that were initially estimated, since the sums indicated do not include the cost for land expropriation and V.A.T. The cost of the double slab foundation system is
compensated to significant extent by the value of the covered parking
spots, each one of them have the size of a large garage box (6 by 3
metre). The number boxes exceed that of the apartments. It is thus reasonable to assume that the real cost of the foundation system is actually a fraction of that indicated in about 30,000 € per apartment. If a
fraction of 30% would be assigned to the foundation itself, the cost per
square metre of the living space should result to be less than 1,400 €,
including foundations: a reasonable value, especially considering the
compressed time of construction, that was made possible by a continuous work over the 24 hours, with three turns of 8 hours each, day and
night and considering as well the very high quality of the buildings for
aspects related to energy consumption, environment and detailing quality.
Parametric costs of the whole intervention, based on 164,29 equivalent buildings, v.a.t. not included
RESEARCH - L’Aquila Earthquake
It is interesting to note that the seismic isolation cost only about 1,5%
of the total, or rather just a bit over 2% when only the building cost is
The modest cost of general and technical activities has to be noted,
made possible by the way the project was managed, extensively
discussed in the previous sections. The real share of the technical
costs of the ForCASE consortium (design, management, security, etc)
has been around 8 million euro, i.e. not much more than 1% of the total
cost. The costs of the furniture includes everything, from TV sets to
bed sheets.
At the moment this article is being completed (September) the last 20
buildings are being constructed, with a significantly lower cost because they are built in already inhabited areas. The decision of the Civil
Protection Department to build additional houses was motivated by
upgraded in the population census. It is foreseen that these buildings
will be delivered within February 2010, with the possibility of notable
anticipation if the meteorological circumstances should be favourable.
The delivery of the houses started Tuesday 29 September with about
500 apartments and will continue at a pace of approximately 300 apartments a week. The property of the buildings will be eventually assigned
to the city of L’Aquila who will be responsible for the management and
maintenance based on pre-defined procedures, specified in detail in the
project documents. Political and economical choices, with relation to
the progress in the repair, strengthening and reconstruction of the buildings damaged in the historical centre and in the city outskirt, will drive
the decision on rent costs and use of the new villages.
The users of the houses are being carefully selected jointly by the city of
L’Aquila and the DPC, taking into account the preferences expressed by
the homeless people, parameters connected to the family situation (number of components, age, economical capacity, etc.) and the localisation of
the original place of living. A prerequisite to be considered is that a
family previous home should have been classified in the category ‘non
easily to be repaired’ (type E and F in the classification of damage).
Forecast of the number of apartments and beds available in function of the foreseen completion
period (forecast of September 22nd, including the twenty buildings that had just been added, of
which forecasts are cautious)
RESEARCH - L’Aquila Earthquake
The houses will anyway become part of the city’s heritage and in the
future it will be therefore possible to reuse them to host vulnerable categories of population (such as the elderly people) or to host students, a
need particularly relevant in L’Aquila, where a significant fraction of
the 25,000 students come from other regions.
In the near future the availability of student housing at a controlled
price could become a relevant peculiarity of the university, modifying
its attraction capacity in a positive way.
Those who have contributed to the success of the C.A.S.E. project:
bani, Marco Vecchietti, Paolo Verri, Stefano Vitalini (site responsible for plates), Roberta Viviani
Department of Civil Protection
Giacomo Aiello (legal responsible),Emilia Aloise, Giovanna Andreozzi, Enrico
Ardito, Vincenzo Ardito, Arianna Bertelli, Mariano Bonvegna, Angelo Borrelli
(administration responsible), Fabrizio Bramerini, Cristina Capriotti, Maria Teresa
Cartolari, Mario Cera, Claudia Ciccone, Pietro Colicchio, Alessandra Conti, Marco
Conti, Maria Laura Conti, Chiara D’angelo, Antonella De Felice, Giulio De Marco,
Giovanni Di Achille, Giovanni Di Mambro, Mauro Dolce (technical responsible),
Riccardo Fabiani, Maria Cristina Ferroni, Salvatore Fiengo, Claudia Fiore,
Mariasilvia Gianneramo, Beatrice Guerra, Gerarda Iannarone, Federica La Chioma,
Luisa Marinaro, Lucia Palermo, Francesca Paneforte, Ada Paolucci, Roberto
Pesolillo, Giancarlo Piccione, Patrizia Picuti, Immacolata Postiglione, Giuseppina
Sementilli, Vincenzo Spaziante (general coordinator), Tiziana Tarduini, Vergilio
Tidei, Fabiola Toni, Angelo Vici.
CASE Consortium
Fabio Aldrovandi, Francesco Ambrosi, Francesco Amici, Maurizio Ardingo (responsible for construction safety measures), Luciano Baglione, Giovanni Bastianini, Paolo
Battegazzore, Giuliano Bellini, Maria Teresa Dolores Bertelegni, Federica Bianchi,
Saverio Bisoni, Gaia Boggioni, Filippo Bonali, Barbara Borzi, Maria Benedetta
Bossi, Matteo Bottari, Vittorio Bozzetto, Roberto Brandimarte, Piero Burba,
Maurizio Calderari, Andrea Caligari, Maura Castellani, Gian Michele Calvi (project
leader, designer and construction director), Salvatore Caroli, Christian Caroli, Paolo
Caroli, Francesco Ceribelli, Antonio Coccia, Andrea Colcuc, Oliviero Comand,
Massimiliano Cordeschi, Filippo Dacarro, Michele D’Adamo, Alberto Damiani
(responsible for building construction), Pietro Damiani, Edi Danielis, Simonetta Di
Nicola, Maurizio De Santis, Pasquale Di Marcantonio, Dante Di Marco, Stefano
D’Ottavio, Ettore Fagà, Mario Fanutti, Carlo Florio, Pierluigi Fontana, Fabrizio
Frau, Renato Fuchs (organisation coordinator), Nicola Gallina, Marco Gasperi,
Fabio Germagnoli, Federico Gianoli, Daniele Gimnetti, Sergio Giordano, Stefano
Grasso, Carlo Lai, Massimo Lardera (responsible for infrastructure), Ignazio Locci,
Giuseppe Lombardi, Mauro Maganetti, Giovanni Magenes, Claudio Maggi, Carlo
Magni, Fabrizio Magni, Michele Magnotti, Gabriele Mantini, Antonio Marcotullio,
Paola Marotta, Sara Martini, Emanuele Meago, Paola Migliazza, Enrico Misale,
Marta Molinari, Federico Monutti, Matteo Moratti (site responsible for structures),
Vincenzo Pane, Vincenzo Paolillo, Alessandro Papale, Carmine Pascale, Pierluigi
Pascale, Moreno Pavan, Fausto Pedetta Peccia, Gianfranco Peressutti, Edoardo
Peronace, Michele Pescina, Paolo Petrucco (site responsible for plates and infrastructure), Piero Petrucco, Nereo Pettenà, Dario Pietra, Roberto Pitolini, Federica
Polidoro, Alessandro Pollini, Stefano Pozzi, Salvatore Provenzano, Bruno Quadrio,
Nadia Rizzardi, Enzo Rizzi, Fabio Roiatti, Cristiana Ruggeri (responsible for mechanical and electrical installations), Gaetano Ruggeri, Mario Rusconi, Daniele
Sambrizzi, Valentina Scenna, Matteo Schena, Michele Schiabel, Paolo Scienza,
Fabiola Sciore, Roberto Scotti, Domenico Sgrò, Martino Signorile, Danilo Marco
Siviero, Luigi Spadaro, Davide Tagliaferri, Piergiuseppe Tamburri, Alessandro Tosello, Stefan Trenkwalder, Roberto Turino (site responsible for buildings), Diego Ur144
Checker team, administrative aspects
Giovanna Andreozzi, Maria Laura Conti, Alessandra Conti, Michele D’adamo, Giovanni Di Mambro, Salvatore Fiengo, Giorgio Grossi, Emilia Aloise, Mariano Bonvenga, Carlo Bordini, Cristina Capriotti, Maria Teresa Cartolari, Carluccio Codeghini, Fabio Compagnoni, Dario Compagnoni, Massimo Criscuolo, Antonella De
Felice, Giuseppe Fasiol, Maria Cristina Ferroni, Arturo Furlan, Achille Gentile,
Alessandro Greco, Gerarda Iannarone, Giuseppe Ianniello, Giovanni Infante, Ettore
Iorio, Paolo Marchesi, Luca Pagani, Lucia Palermo, Roberto Pesolillo, Salvatore
Provenzano, Rosario Romano, Gianni Strazzullo, Fabiola Toni, Daniela Ursino,
Michele Villani
Checker team, structures
Edoardo Cosenza, Gaetano Manfredi, Claudio Moroni, Paolo Pinto (President),
Paolo Zanon (assistenti: Massimo Acanfora, Claudio D’Ambra, Antimo Fiorillo)
Companies and organisations
Excavations: CO.GE.FER. s.p.a.; Midal s.r.l.; P.R.S. Produzione e Servizi s.r.l.
Concrete: Colabeton s.p.a.; Società Meridionale Inerti SMI s.r.l. Steel reinforcement:
La Veneta Reti s.p.a. Steel columns: A.T.I. Edimo Metallo s.p.a. /Taddei s.p.a.;
Cordioli & C. s.p.a.; Formwork and assistance: Consorzio Edile C.M. Gruppo Bison;
Sacaim s.p.a.; Zoppoli & Pulcher s.p.a. Isolators: Alga s.p.a.; FIP Industriale s.p.a.
Buildings: A.T.I. Consorzio Stabile CONSTA s.c.p.a./Sicap s.p.a.; A.T.I. Donati
s.p.a./Tirrena Lavori s.r.l./Dema Costruzioni s.r.l./Q5 s.r.l.; A.T.I. Eschilo Uno
s.r.l./COGEIM s.p.a./Alfa Costruzioni 2008 s.r.l.; A.T.I. Ille prefabbricati
s.p.a./Belwood s.r.l.; A.T.I. Impresa Costruzioni Giuseppe Maltauro s.p.a./Taddei
s.p.a.; A.T.I. Iter Gestione e Appalti s.p.a./Sled s.p.a./Vitale Costruzioni s.p.a.;
A.T.I. COGE Costruzioni Generali s.p.a. /Consorzio Esi; Consorzio Etruria s.c.a.r.l.;
Consorzio Stabile Arcale; Cosbau s.p.a.; D’Agostino Angelo Antonio Costruzioni
Generali s.r.l.; Impresa di Costruzioni Ing. Raffaello Pellegrini s.r.l.; Meraviglia
s.p.a.; Orceana Costruzioni s.p.a.; R.T.I. Ing. Armido Frezza s.r.l./Walter Frezza
Costruzioni s.r.l./ Archilegno s.r.l.; Wood Beton s.p.a. Furniture: Del Tongo
Industrie s.p.a.; Estel Office s.p.a.; Mobilificio Florida s.r.l.; R.T.I. Europeo s.p.a./
PM.International Furnishing s.r.l. Infrastructure: CO.M.AB. Appalti Pubblici e
Privati s.n.c.; Codimar s.r.l..; Codisab s.r.l.; Conglomerati Bituminosi s.r.l.;
Facciolini s.r.l.; G.C.G. s.r.l.; I Platani s.r.l.; Impresa Edile Di Cola Michele; Ing.
Armido Frezza s.r.l.; Molisana Inerti Conglomerati s.r.l.; Produzione e Servizi s.r.l.;
Ridolfi Idio e Figli s.r.l.; San Giovanni Inerti di Pietro Mascitti s.r.l.; Valentini
Costruzioni s.a.s.; Elevators: Marrocco elevators s.r.l., ATI S.A.S. s.r.l./Grivan Group
s.r.l., Schindler s.p.a.; Photovoltaic panels: R.T.I. Ener Point s.p.a./Ener Point
Energy s.r.l./Troiani & Ciarrocchi s.r.l.; Green areas: R.T.I. 3a Progetti/Gsa
s.r.l./O.Ci.Ma. s.r.l./Bellomia-Sebastianini-Euroengineering s.r.l., Consorzio
Sestante. Demolition: CODISAB SRL, A.S.M. s.p.a.; Connections to external pipeline networks: ENEL Rete Gas, ENEL Energia, GranSasso Acqua.1
RESEARCH - L’Aquila Earthquake
1. NTC (2008) - Norme Tecniche per le Costruzioni, D.M. 14/01/2008, Gazzetta Ufficiale 04/02/2008, Italia.
2. Zayas V., Low S. (1990) - A Simple Pendulum Technique for Achieving Seismic
Isolation, Earthquake Spectra, Vol. 6, No. 2.
3. Almazan J.L., De la Llera J.C. (2002) - Analytical model of structures with frictional pendulum isolators, Earthquake engineering and structural dynamics, Vol.
31, 305-332.
4. Calvi G.M., Ceresa P., Casarotti C., Bolognini D., Auricchio F. (2004) - Effects of
axial force variation on the seismic response of bridges isolated with friction pendulum systems, Journal of Earthquake Engineering, Vol. 8, SI1, 187-224.
5. Christopoulos C., Filiatrault A. (2006) - Principles of Passive Supplemental
Damping and Seismic Isolation, IUSS Press, Pavia.
6. Priestley M.J.N., Calvi G.M. (2002) - Strategies for repair and seismic upgrading
of Bolu Viaduct 1, Turkey, Journal of Earthquake Engineering, Vol. 6, SI1, 157-184.
7. Tsai C.S. (1997) - Finite element formulations for friction Pendulum seismic iso-
lation bearings, International Jour. for Num. Methods in Engineering, Vol. 40,
8. Wang Y., Chung L.L., Liao W.H. (1998) - Seismic response analysis of bridges
isolated with friction pendulum bearings, Earthquake engineering and structural
dynamics, 27, 1069-1093.
9. Priestley M.J.N., Calvi G.M., Kowalsky M.J. (2007) - Displacement based design
of structures, IUSS Press, Pavia.
10. Crowley, H. and Pinho R. (2004) - Period-height relationship ofr existing
European reinforced concrete buildings, Journal of Earthquake Engineering, Vol. 8
(SP1), 93-120.
11. Crowley H., Stucchi M., Meletti C., Calvi G.M., Pacor F. (2009) - Uno sguardo
agli spettri delle NTC08 in relazione al terremoto de L’Aquila, capitolo 1.7 in questo volume.
12. AA.VV. (2007) - Definizione dell’input sismico sulla base degli spostamenti,
progetto S5 INGVDPC, http://progettos5.stru.polimi.it.
13. Comité Européen de Normalisation, Eurocode 8 part 2 (2006) - prEN1998-2,
CEN, Brussels.
Here and in the next pages, some pictures of the completed buildings areas.
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
RESEARCH - L’Aquila Earthquake
Italian High-Speed Network
A special focus on concrete structures
he new Italian high-speed Network, due to the needs of increasing the
capacity of the actual railway operating lines, nearly doubling the
actual number of trains running daily, and decreasing travel time, has involved in the last fifteen years, large engineering resources, construction skills
and strict planning, managing organization, railway engineers supervision,
apart from the necessary economical huge investments.
The main characteristics of the new lines and the strategic choices from the
infrastructural and structural points of view are presented in the next chapter.
The new lines are designed according to the national code for the design
and construction of railways bridges [1], and to the most advanced technological standards in order to achieve the safest conditions of service, speed
and interoperability with operating railways and with the European highspeed lines of railway transportation of passengers and freight.
More than 90 % of bridges and viaducts of the new high-speed lines are
realised with simply supported spans of prestressed concrete (PC) decks,
so attention will be paid on their typical structural solutions.
Simply supported composite steel and concrete spans, few continuous bridges and some special structures (arch bridges and the cable-stayed bridge
over Po River [2]) compose the remaining part.
Main concepts of durability of concrete structures required in the design of
railway bridges are then focused.
Construction processes are then described, taking examples from the latest viaducts with PC decks.
Finally, main concepts in the design of continuous PC bridges are outlined.
Fig. 1- Italian New High-speed Network: lines under design, construction and the operating line Roma-Firenze, built in the
70’s-80’s for a design speed of 250 km/h [3].
Strategic choices
Traffic analysis carried out on existing Italian railway network at the end of
the 1980 remarked the following needs: quadrupling the main passenger
transport routes, upgrading and increasing freight transport, reducing time
of travel for passengers trains, integrating Italian network with European
So, the first infrastructural choice was the realization of a new mixed passenger/freight high-speed network with a close integration with existing
lines and with interchange centres (interports, ports, airports).
The close integration with the existing conventional network will produce an
increase of freight transport capacity on the “historical” lines, clearing the
existing network from the long distance passengers traffic, and an increase
of freight traffic using new lines during specific time bands (usually at night).
The choice of a mixed traffic meant low ruling gradients (less than 12 ‰)
and heavy design loads (SW0/SW2) adopted in the new Italian standard
for railway bridges [1]. For this reason, the standard was rewritten in 1995,
then revised in 1997.The old standards had been written 50 years before
and were related only to conventional lines.The Italian standard for railway
bridges [1] has introduced LM71 (passengers traffic as showed in leaflets
UIC 702 and 776-1), SW/0 and SW/2 (heavy traffic) models of loads
according to ENV 1991-3: Actions on structures, Part 3:Traffic loads on bridges (Ed. 1996).
These new standards are in perfect agreement with the European Technical
Standards for Interoperability of the trains in the European High-speed
One of the main prescriptions asks for a structural design respectful to all
prescriptions for seismic areas (at least III category – the minimum considered in the 1996 Italian seismic code), even in no-seismic areas, apart
from Sardinia. A proper standard was written and recently revised for the
design of railway bridges to be built in seismic area [4]. It deserves to be
pointed out that many no-seismic areas became recently seismic, in the
latest proposals of codes, giving interesting confirmation to the conservative railways code approach.
According to general seismic design principles, the adoption of special rules
and technical details is requested to guarantee a minimum ductility of the
structure, and it has direct consequences on the care for the details of
reinforcement design of piers and foundations.
Two main characteristics of the applied national code [1] are the concepts
of train-track-structure dynamic interaction and train-rail-structure static
The dynamic interaction analysis is evaluated to check the safety of the
train and the comfort of the passengers, with an analysis of the following
parameters: all decks for high-speed railway must respect the limit value of
2.5 as maximum dynamic amplification of static deflection (“impact factor”
j real=j dyn /d stat) and the value of the vertical acceleration at deck midspan, induced by real trains running at different speed (from 10 km/h up
to 1.2 maximum speed of the line), must be lower than 3.5 m/s2.
For standard simply supported beam or continuous bridges with total
length shorter than 130 m, a simplified analysis can be adopted according
CONSTRUCTION - Civil engineering Works
to Annex A of [1], with a preliminary check of flexural - frequencies of vibration modes: the simply supported prestressed concrete decks always
respect the limit values specified in [1], being first flexural mode frequency
between 4 and 8 Hz.
For non-conventional structures, as arch bridges, cable-stayed bridges, etc.,
a “Runnability” analysis is required.The analysis has to consider all dynamic
characteristics of the system: railway structures, suspension system of the
vehicles, rail fasten system etc., track and wheel irregularities.
The static interaction analysis studies the effects on rail and bridge structure due to variation of thermal conditions in the structures, to the longitudinal forces associated to braking and traction, and to the longitudinal
displacements due to vertical loads. For simply supported prestressed concrete decks, which respect a maximum length of 65 m and small variations
of the longitudinal stiffness of piers and foundations, a simplified method
can be adopted as indicated in Annex B of [1].
In any other case, it is necessary to analyse advanced Finite Element models
to evaluate these effects. The analysis has to check rail stress limits with
maximum value of compressive stress 60 MPa and maximum value of tensile stress 70 MPa, the relative displacement between deck bridge and the
rail, and the forces acting on the bearings.
Also reliability under service conditions is required: comfort limit state has
to be verified for a maximum midspan deflection with the load of one
LM71 load model, increased with dynamic factor. This deflection must not
exceed l/2400 for design length of the span l<30m, l/2800 for
30m<l<60m and l/3000 for l >60m. Maximum deformability of structures under train load is checked to keep the contact rail-wheel safe and stable: deck torsion, rotation at supports and horizontal deflection have to be
evaluated. The limits of deformation of the structures are similar to those
pointed out in the same Eurocode, and are widely respected by common
simply supported spans.
Special attention has been put in the concepts of durability of structures for
railway bridges, introduced in [1]. As the subject deserves wide illustration
and details, a full paragraph has been devoted to the scope.
As high-speed network is designed for the use of long welded rail, the structural system for the viaducts must avoid rail expansion devices. In all highspeed network, only along the Milano-Bologna line, for the crossing over Po
River, composed by two continuous bridges and the cable stayed bridge [2],
two joints in the rails have been necessary to keep the expansion length
within allowable limits.
Other rules and prescriptions for design and construction are taken into
consideration in [1] and are illustrated in the following paragraphs. All these
are finalised to have low costs of maintenance of the infrastructure, to minimise the irregularities of the track and to reach a high performance level
in the field of the durability and reliability of the system.
Simply supported prestressed concrete bridges
Simply supported spans of prestressed concrete deck realised more than
90% of the new lines. It is undoubtedly a traditional choice of Italian Railway Company (Ferrovie dello Stato – FS) for ordinary viaducts: to better fit
CONSTRUCTION - Civil engineering Works
with long welded rail, to avoid rail expansion devices, to ease maintenance
operations and minimize maintenance costs. Besides, this solution is usually preferred in those cases when bridges have to be designed in areas with
compressible soils or in river channels.
Figures and statistics of this paper are based on more than 600 km long
high-speed lines. Attention will be paid to double-track decks, with a distance between tracks of 5.0 m, designed for a train-speed of 300 km/h, and
for both heavy and passengers traffic load models [1], with rails on prestressed concrete sleepers on ballast.They count nearly 2300 spans of simply supported prestressed concrete bridges, and they are composed of
nearly 6400 precast beams or monolithic decks.
In order to reach this frame of prestressed concrete decks, analysis will concern the structural characteristics such as use of pre-casting, tensioning
systems, bearings, expansion joints, all durability issues as multi-layer protection systems, monitoring, methods of construction and costs.
Deck The pre-stressed concrete elements are realised with both pre-tensioning and post-tensioning systems.The post-tensioning systems are always
designed with bonded internal cables, even if Italian standard for railway
bridges [1], generally speaking and under severe controls, allows also external post-tensioning. According to [1], post-tensioning cables composed by
bars should be preferred for viaducts along railways with electric traction of
direct current, and both solutions with cables composed by strands or bars
can be used in structures for railways with electric traction of alternating
current as the high-speed network.
In [1], special attention for durability and limiting or avoiding cracking of
concrete is introduced: undoubtedly most limiting verifications deal with limitation of maximum compressive stresses and, especially, strong limitation of
tensile stresses during construction and final conditions. In particular, no longitudinal tensile stress in PC structures is admitted, with maximum design
loads and both Allowable Stress or Limit States methods of verification.
Besides, cracking of concrete must be verified towards no decompression
limit state for verification under track equipment, where inspection is not
The experience of existing railway lines with concrete structures with possible beginning of corrosion of the reinforcement and spalling of the concrete, which leads to easier access to the pre-stressing tendons for aggressive agents, has been translated into design prescriptions.The required concrete cover to reinforcement, tendons and pre-tensioned strands has been
increased, compared to Italian standard for design of structures. Minimum
concrete cover thickness is required to be 3 cm for PC decks, increased to
3.5 cm under track equipment, at least one external diameter of duct in
case of post-tensioning, and 3 strand diameters in case of pre-tensioning.
Mix design of concrete for PC deck has to respect a 0.45 water to concrete
ratio, a S4÷S5 concrete consistency class of at least 45 MPa characteristic
cubic strength. A quality assurance system and testing before and during
every casting operation reveals the quality of mix design, which is recognised as an important factor for life and durability of PC structures.
Bearings and expansion joints Under simply supported railway bridges, only one kind of bearing is generally present: spherical bearings with
polished stainless steel and PTFE plate.
Italian High-Speed Network
With this kind of bearing, rotations can occur till ±0.0167 rad in all directions, in order to place the bearing without inserting packings. In order to
avoid parasite forces arising with only one train on a double line bridge
deck, a new kind of fixed bearings has been studied: it has a special device
which controls horizontal stiffness.
The cover of expansion joints are realised with dielectrical elastomeric
cushion joints, composed by neoprene reinforced with vulcanized steel plates.They allow fast bearings changing with a maximum differential lifting of
50 mm between decks, operated by hydraulic jacks between deck and pier
cap (all simply supported or continuous decks have to pass this design verification) without any operation under rails. Actually, the use of mechanical
devices instead of resins avoids disease to daily train in case this operation
becomes necessary.
On pier caps and abutments of every bridge span, reinforced concrete or
steel devices (“stroke end device”) are required in order to avoid deck slipping out of pier cap and falling, because of accidental breaking of fixed bearings e.g. in case of devastating earthquakes.There are pillows of reinforced
neoprene where decks may hurt against these provisions and their maintenance or changing operations has to be assured by proper design.
Italian standard for railway bridge bearings and cover of expansion joints
requires these devices underpass preliminary homologation tests led by F.S.
technicians, through prototypes testing, in order to assure quality of every
single component and of the final assembled products.
Piers and foundation Piers have usually circular or rectangular, full or
empty, cross-sections, while foundations are usually realised with plinths with
large diameter reinforced concrete piles.
In case of piers in riverbed, even if empty structural sections are adopted,
low class of concrete is always poured inside till the river maximum level, in
order to avoid unexpected water inside.
As previously mentioned, in order to increase structural safety, all bridges are
designed considering at least low seismic condition: it focuses the designers'
attention especially on reinforcement details, very important for piers and
piles. Good number of stirrups and loops for longitudinal bars and concrete confinement, use of hooks for good stirrup behaviour, limitation of maximum compression stress in pier concrete, no junction or superposition of
longitudinal bars in the length of 3 m from foundation, etc. are consequences of above-mentioned prescriptions.
The minimum reinforcement areas for both piles and piers is fixed to the
0.6 % area of concrete section, and spirals are admitted as stirrups in
reinforced concrete piles only if welded to longitudinal bars in every intersection.
Type a
Weight of one precast box: 455 ton (33.1 m)
19% of total length of viaducts
915 ton one deck weight (34.5 m)
343.196 ton is total weight
Type b
Weight of single precast V beam: 88 ton
40.5% of total length of viaducts
650 ton one deck weight (25 m)
698.476 ton is total weight
Fig. 2- Two box girders deck (a) and four precast V beams and cast in situ slab (b).
beams with post-tensioning cables. In Roma-Napoli line it was also realised
with two V-beams and cast in situ slab. Transversal beams are usually prestressed with straight cables of strands or bars.
The number of transversal beams is prescribed in [1]: for a deck with two
or more girders, at least two prestressed concrete transversal beams have
to be designed out from supports and more in case of decks longer than
25 m.
Strands getting out from the heads of the box girder are cut, isolated and
protected with the use of dielectric resin. Decks' deformability is largely verified for comfort limit state: maximum deflection at midspan for Type “a” is
less than l/5600.
Type “b” is composed by four precast V-beams and cast in situ slab: actual
maximum length is 33.6 m.
Beams are steam cured, pre-tensioned with longitudinal steel strands and
Typical cross sections
The most common cross-sections of prestressed concrete decks are showed
in Fig. 2, 6 and 11; in the following, a brief description of main features is
presented for each typical cross-section.
Type “a” is a box girders deck, spanning till 34.5 m, generally composed by
two precast box girders, prestressed with longitudinal steel strands and connected with small second step casting in the slab and with transversal
Fig. 3- Prestressed concrete V beam (type b) in stocking area, Torino-Milano line.
CONSTRUCTION - Civil engineering Works
transversally connected with cables in transversal concrete beams; it is the
most common deck: it has been chosen for 40.5 % of total length of simply supported prestressed concrete deck.
The maximum deflection at midspan is largely verified: for type “b” spanning 25 m (22.3 % of length of all prestressed concrete decks) it is less
than l/6000.
Because of the prescription of complete absence of tensile stress during
construction and life of the bridge, and of the large amount of pre-tensioning strands in V-beams, the technique of strand passivation for few metres
along beam ends, over supports, has been introduced for a portion of
strands, in order to reduce even minimal cracking on heads of the beam.
From an aesthetic point of view, shortest spans of box girders (both V-beam
or cellular deck) can be put at a disadvantage, because in [1] a free height of at least 1.6-1.8 m inside box girder is prescribed to be left to ease
inspection, leading to relevant height of the deck even for short span bridge. Anyway, these spans can be agreeably inserted in case of viaducts with
short piers.
Type “c” is a single box girder deck, realised in two different ways and
Fig. 5- The first precast 25m long single box girder (Torino-Milano), during launching operations.
sversal beams or second step castings of concrete in head anchorages of
tendons in the required transversal beams, which always become visible
with time.
Anyway, in some case of grillage deck, when perspective had to be improved, concrete noise barriers have been usefully adopted, covering second
step casting or empty spaces on pier-cap between decks for inspection.
Type “e” is composed by four precast I beams and cast in situ slab.
Beams are longitudinal post-tensioned with cables composed by strands
with straight and parabolic profiles.
Some are tensioned in precasting plant, then, after completing the bottom
slab and tensioning of transversal cables, the second part of longitudinal
cables is tensioned over the piers and slab is casted. Longest span of type
“e” is also the longest span for simply supported prestressed concrete
decks: 46.2 m.
Type “e” has the advantage to manage precasting and launching of one
beam at time instead of full deck, so requiring simpler technology, but, at
the same time, the operation of assembling formworks and casting connection of lower slab and transversal beams, and the huge transversal posttensioning (no. 47 4-strands cables for 46.2 m long deck) may put it to a
Type “f” is the original Modena viaduct: the first case of lower way U deck
for high-speed lines; the double track is realized by two independent decks,
piers and common foundation. It has two single track decks spanning 31.5
m, and a total width of 18.4 m; each deck is pre-stressed with 20 longitudinal post-tensioning tendons of 12 strands. 566 km of corrugated plastic
Type c
Weight of single precast deck: 567 ton (25 m)
Weight of cast in situ box deck: 1043 ton
(43.2 m)
11% of total length of viaducts
173.856 ton is total weight
Fig. 4- Box girder (type a) on carriers towards launching operations, Milano-Bologna line.
lengths: 25 m long precast box girder with longitudinal pre-tensioned steel
strands on Torino-Milano line (3.78 km long Santhià and 1 km long Carisio
viaducts) and cast in situ post-tensioned deck spanning 43.2 m (2.8 km
long Padulicella viaduct) on Roma-Napoli line.
Type “d” is adopted in 5.1 km long Piacenza viaduct: 150 precast spans
with two cells and curved transversal profiles. It is a single monolithic box
girder of 970 ton, with a maximum length of 33.1 m.
Piacenza viaduct has been provided with 119 km of corrugated plastic
ducts and it represents the first application of electrically isolated disposals
for the anchorages of 12 and 19 strands for longitudinal post-tensioning
cables. Compared to the grillage decks, the monolithic decks have the
aesthetic advantage of clean prospects and even deck sides, without tranCONSTRUCTION - Civil engineering Works
Type d
Weight of single precast deck: 970 ton
7.2% of total length of viaducts
145.500 ton is total weight
Fig. 6- Box girder with single cell (c) and Box girder with two cells (d).
Italian High-Speed Network
Fig. 7- Piacenza viaduct precast deck, 33.1 m, in stocking area.
Fig. 8- Perspective of curved profiles of Piacenza viaduct.
Fig. 9- Four precast I beams in stocking area, from Roma-Napoli high-speed line.
Fig. 10- Four precast I beams launched over pier caps, from Milano-Bologna high-speed line.
Type e
Weight of single precast I beam:
270 ton (46.2 m)
Weight of four I beams deck:
1.400 ton (46.2 m)
7.6% of total length of viaducts
137.274 ton is total weight
Type f
Weight of single precast
deck: 689 ton
15% of total length of
446.472 ton is total
Fig. 11- Four precast I beams and cast in situ slab (e) and Lower way U deck (f).
Fig. 12- The first one of 750 Modena precast decks in stocking area, Milano-Bologna line.
Fig. 13- Beam head of Modena precast deck in stocking area, Milano-Bologna line.
ducts is used. In Milano-Bologna line this structural solution for prestressed
concrete deck is used for 10.7 km long double track viaducts. It is also used
in Junctions, for another 3.33 km of single-track, for a total number of 767
precast spans. In particular, Modena viaduct is the longest viaduct of all
high-speed lines, with its length of 7.1 km.
The structural solution of lower way deck has been usefully adopted to
minimize structural height plus noise barrier, because the noise barriers
became a part of the structure. Modena system of viaducts is one of the
few cases, where aesthetics and environment impact have so strongly led
the process of design and the solutions for structural and construction
needs, in order to have quite original deck, unique in its kind.
Through periodical inspection of old railway lines and particularly of the
Roma-Firenze railway line, an analysis of the most common defects on bridge deck have been made in order to work out new guidelines of design and
construction of the high-speed lines. Inadequate access for inspection,
leaking of waterproofing system of the expansion joints between decks,
insufficient cover, not efficient bearings towards maintenance operations as
lifting of decks in order to change bearings, or reduced space to insert
hydraulic jacks on pier-caps, etc. have been found as most frequent defects.
These circumstances led to the introduction of new or more detailed prescriptions in the Italian standard for railway bridges [1], in order to guarantee better behaviour of structures with time, care for durability of concrete structures, tensioning cables, system of waterproofing, devices for good
drainage and anything else whose purpose is to ensure the overall longterm integrity of bridge structure. But, first of all, great care is put to inspection ways to check the conditions of the structures and medium-life elements.
Access for inspection All bridges are designed assuring access for
inspection, testing, maintenance and possible replacement of medium life
elements. It must be always possible to walk over bridge decks because a
width of min. 50 cm on both sides is left for maintenance people. Inside cellular deck or closed box girders, as previously mentioned, a minimum height of 1.6-1.8 m must be always guaranteed; fixed stairs from deck to pier
cap must be provided every 3 spans or 100 m and from piers to the
ground every 500 m, for viaducts longer than 1000 m. Over pier cap it
must be always possible to pass from one deck to the following and stairs
or landings are fixed to ease the movements of maintenance people.
Finally, it must be also possible to inspect bearings and stroke end devices
or to operate in case of replacement of bearings or neoprene pillows, so a
free height of 40 cm is left between lower side deck and top of pier-cap.
Drainage It is a key issue about durability; deck slabs are provided with
provisions for good drainage and great attention is put to design, testing and
layout of all devices. Drainage of expansion joints is assured by a flashing
tray of elastomeric material stuck with resins to slabs’ ends, in order to
avoid leaking over piers and to drain water out of deck sides. Over bridge
deck, in order to protect from atmospheric agents, a thick layer of waterproofing is extended, also beneath the footways. In case of sensible prestressing system, pre-stressing strands or post-tensioning cable system just
beneath deck slab or critical drainage system (types “a”, “c”, “d” and “f”),
a sprayed polyurethane waterproofing of 3 to 5 mm is extended. Checks
are been carried out on site for adhesion and thickness by F.S. technicians.
This surface treatment has proven to be very long life cycle performant.
In post-tensioned concrete structures, deck anchorages are to be avoided
on deck slab; anyway, for all anchorages, design has to avoid leakage to get
access to anchorages, providing protection against leaking expansion joints,
as water drips.
Post-tensioning tendonsThe grouting of the sheaths of prestressed concrete bridges is always done with vacuum technique with a depression of
0.2 bar during injection. It is standard for prestressed concrete deck with
post-tensioning tendons because grouting has been recognised as a key
CONSTRUCTION - Civil engineering Works
operation to ensure durability of prestressing. Usually, it is difficult to ensure complete filling of the ducts: pathology teaches us that the prestressing
tendons are more vulnerable in the case of post-tensioning than in the case
of pre-tensioning.
Vacuum injection should avoid having air bubbles near high points of tendon profile and in case of simply supported structures high points coincide
always with anchorage locations. Besides, all end caps are filled with grout
and surrounded by concrete held in place by reinforcement, with noshrinking concrete and the same compressive strength as deck concrete.
Latest tendencies about durability of anchorages are High Density
Polyethylene ducts with plastic end cap left in concrete and the use of electrical isolated tendons as a further protection of the tendon and mean for
monitoring: in Milano-Bologna and Torino-Milano railway lines, a large scale
application of plastic ducts and electrical isolated tendons has been undertaken. Italferr has taken the Technical Report from fib (fib 2000) and Swiss
guidelines on electric isolated tendons (2001) as standards, asking for
mechanical and chemical tests, field measurements to be undergone; a
System Approval testing has to be conducted on site on every first application of a prestressing system.
Deck equipment Every bridge deck is furnished of railings on both sides,
anchorages of noise barriers for their future assembly, electric traction
poles, stairs to pier cap and from piers to the ground. Great attention has
been put to deck equipment: every steel finishing is installed with linkages
electrically isolated from deck reinforcement and connected to a dissipative end in the earth for safety reasons. Stainless steel is preferred for the
most sensible connections.
Electrical isolation of structures Every bridge deck is electrically isolated through isolated bearings and expansion joints, from piers and the
other decks, besides, in order to prevent and protect bridge reinforcement
against strain currents, few disposal for every deck are disposed in an
accessible area in order to measure strain current and isolation grade after
traffic activation.
This is probably a minor problem on high-speed line decks, but felt deeply
in every normal bridge deck and in the Junctions of high-speed line.
Whenever problems of potential differences should arise, structures will be
electrically connected to earth or a cathodic protection should be eventually adopted.
In case of post-tensioned structures, all anchorages are electrically connected (when no electrical isolated tendons is adopted) and the terminal is
drawn out of the structure in order to provide eventually in the future the
same provisions as for deck reinforcement, otherwise, in case of pre-tensioned decks, the head faces of the beam are protected with synthetic dielectrical resins.
Monitoring and Maintenance In order to improve deck’s behaviour
knowledge and control it with time under the influence of external agents
(environmental actions, traffic loads, seismic events or exceptional hydrogeological events), a complex monitoring system integrated with the highspeed line has been designed.
At least one section (deck, pier, foundation, piles) per viaduct and, in case of
long viaducts, one section every 1000 m is instrumented: it means a large
CONSTRUCTION - Civil engineering Works
number of strain-gages, inclinometers, thermocouples, instrumented bearings, load cells, foundation settlement meters, piezometers etc. Seldom
accelerometers are provided in order to evaluate dynamic response of the
structures also in case of seismic actions.
Maintenance program of high-speed lines is essentially based on maintenance actions followings inspection visits: in the code 44/c of Italian railways
about lines maintenance [5], frequencies, ways of inspection and following
check schedules are prescribed. These check schedules have the double
aim to check the safety of structures towards train traffic, and to keep
memory of time evolution of the behaviour of structures. According to [5]
every year a program of action has to be adopted to eliminate anomalies
encountered in structures or to face critical situations.
Construction process
As all bridge designers know, construction process may deeply influence
design choices, also in case of pre-cast prestressed concrete beams, which
compose the majority of our new bridges and viaducts: the construction
scheduling, the technologies of lifting, transporting and lowering over the
piers are investigated.
To improve the overall quality of the infrastructure design and production,
a quality control system is implemented during both the design phase and
the construction phase. In the following, main construction features of each
structural solution are described in the text and representative construction
phases are showed in the pictures.
Precast beams and cast in situ slab When talking about construction
features we must divide our decks in few major families: one of these is the
precast V or I beams with cast in situ slab.The short spans of four V-beams
deck of type “b” (Fig. 14-15) and the shortest I-beams of deck of type "e",
of an average load of 100 tons each, are the only ones which can be casted
and pre-tensioned in pre-casting plant, moved on ordinary roads and led on
site, where each beam is lifted to its final position.
Then predalles or formworks are assembled, reinforcement is laid and the
concrete slab is casted over the piers. As it is not necessary to pass over
completed decks, ordinary roads can be used and there’s no obliged
sequence of spans’ layout, this method of construction has the important
property of flexibility; besides, relatively simple technology is necessary for
its realization.
Cast in situ box girder In the only case of single box girder Padulicella
viaduct (Fig. 16), the deck is casted and pretensioned over the piers on selflaunching formworks and special casting equipment is used.
To accelerate the production, the reinforcement cage had to be pre-assembled, transported and lowered in the formwork before casting.
Two box girdersTwo precast box girders, each weighting about 450 tons,
are precast and prestressed with pre-tensioned strands, then lifted over the
viaduct where they are transported by two small carriers on tyres, towards
launching operations (Fig.17).
Deck is completed over piers, with second step casting in central slab and
in transversal beams, then transversal strands cables are tensioned and
grouted. The case of two precast box girders is half way between the pre159
Italian High-Speed Network
Fig. 14- Precast V beam lifted up from stocking area.
Fig. 16- Reinforcement cage transported in the formwork over the piers, Padulicella viaduct, Roma-Napoli.
Fig. 15- Precast V beam towards final position over the piers.
cast beams and the full-span pre-casting.
Actually launching operations similar to those of full span pre-casting are
necessary, but second step casting and transversal prestessing in transversal beams are needed.
As the full span precast decks of the following chapter, these box girders
are realized in a plant near the viaduct: all these plants are dismantled at
the end of the works.
Full span precasting In those cases when much more spans had to be
built in short time, full span pre-casting has been preferred (Figs. 18-21).
According to this process of construction, decks are totally pre-casted; no
post-tensioned transversal beam is needed to complete the deck over piers.
Afterwards, one by one they are moved towards launching operations.
A «carrier» and a «support beam» always compose the launching system.
The carrier slowly moves on tyres or steel wheels, lifts a stored beam, and
transports it along the viaduct, moving on the placed girders.
The device for transport forms an integral part of the device for the launching of the girders: the carrier drives then into a second steel girder called
support beam, suspends the beam over its final position.
The support beam is drawn back and the beam is lowered. With four or
six bearings, hydraulic jacks or load cells are used in order to check weight
load distribution, and then the beam is lowered on its final bearings.
Fig. 17. Launching operations of a box girder of S. Rocco al Porto 2 viaduct, on Milano-Bologna line.
Fig. 18- Modena precast deck lifted and transported on tyres from stocking area, Milano-Bologna line.
The use of pre-assembled reinforcement cages, independent casting lines
and several formworks, different phases of tensioning operations and the
use of storage areas for cables’ injection can speed up the process.
On the other hand, if for any reason one deck has to be stopped in the
stocking area, the process may stop for days. So this method doesn’t have
the flexibility pointed out before for beams and slab decks.
CONSTRUCTION - Civil engineering Works
Fig. 22- Concrete, Reinforcement and Strands loads of different typologies of deck versus span length, (figures based on highspeed lines: Roma-Napoli; Bologna-Firenze; Milano-Bologna; Torino-Milano).
Fig. 19- Piacenza precast transported on steel wheels from stocking area towards launching, Milano-Bologna line.
Fig. 20- Modena precast lower way U deck during launching operations, on Milano-Bologna line.
and reinforcement and prestressing steel amounts for each span, apart
from few exceptions.
Talking about deck load, first exception to be mentioned is Modena deck:
as a single way deck it results heavy solution for a double line, but, at the
same time, the simple “U” profile, easy to manage from a design point of
view, doesn't cause similar examples of exception for the load of reinforcement and prestressing steel. Decks composed by beams can never minimize the use of steel because of their transversal connections, while single box
girders seem, even if based on few examples, to behave more efficiently.
Anyway, many other factors are to be considered in the choice of the best
structural and technological solution for a new railway bridge deck: they
may depend on construction method, workmanship factors, number of
spans to be built and required time scheduling of construction, as previously mentioned.
Another factor for evaluating the solution for a bridge deck is, of course, its
aesthetic impact: even if mentioned at the end of the analysis, in some case
it has strictly led the design choices.
Generally speaking, the span length over deck height ratio can be considered one of the simplest measure to evaluate the grade of slenderness: in
case of simply supported PC decks, it ranges between 9 and 12; even for
short spans, the prescription of minimum free height inside cellular decks
causes small ratio.
Anyway considering that the average piers’ height is not more than 7÷8 m,
usually long spans are not used with very short piers because of aesthetic
reasons too.
Monolithic decks are to be preferred to precast beams decks because transversal beams are always impacting on the sides' prospects.
Finally, a good example of agreeable bridge aesthetics in the field of simply
Fig. 21- Santhià precast single box deck during launching operations, Torino Milano line.
Peak cycle is variable: Modena viaduct has a casting and launching speed
of two precast girders per day. Others, as Piacenza viaduct, have the design of the spans and of the casting yard centred on a target peak cycle of
two double-track decks per week.
Comparison between a-f typologies
Last data about PC simply supported spans are in Figg. 22 and 23 where
the most important figures about double-track decks of the new high-speed
lines are presented.
There’s good uniformity between different typologies dealing with deck load
CONSTRUCTION - Civil engineering Works
Fig. 23- Span length distribution (e) among the a-f typologies (figures based on high-speed lines: Roma-Napoli; BolognaFirenze; Milano-Bologna; Torino-Milano).
Italian High-Speed Network
supported spans was obtained changing the structural solution: the Savena
arch bridge on Bologna-Firenze high-speed line, designed to overpass
Savena River. It is composed by reinforced concrete arch, steel hangers connected by spherical hinges to the 2930 ton prestressed concrete deck, with
both transversal and longitudinal post-tensioning cables. It has the lowest
height under rails (1.60 m) for a span length of 62.5 m.
Continuous prestressed concrete bridges
In few cases, for riverbed or embankment crossings, highway or railway
over-flyings, multi-span continuous PC bridges have been designed.The most
common choice for a single long span in a sequence of shorter simply supported spans is the composite steel and concrete deck, from about 40 m
to more than 70 m, but they result often strongly impacting with longer PC
viaducts perspective, consisting generally of a single exception, with different
structural height, colour and side profile.
In other case, the need to harmonize approach viaduct spans length to a
bigger structure as arch bridges or cable-stayed bridge has led to multi-span
PC viaduct with spans of 60÷70 m.
According to ref. [1], all previously mentioned design and durability prescriptions have to be applied to continuous bridge: access for inspection of
every pier cap is more stringent and difficult to obtain, leading often to complex systems of stairs and landings around central supports. Besides, every
deck has to be verified for lifting in case of bearings' replacement and it
may result structurally demanding, while, from a technological point of view,
it can require a specific design to give disposals for 20% additional prestressing in each span longer than 40 m.
PC continuous beams are often casted in phases, it is quite rare to have a
single casting operation for two or three spans and in [1] precast segmental construction is forbidden. For cast in situ segmental PC bridge, a minimum reinforcement area through every joint of 3.0% area of concrete cross
section of deck is prescribed and minimum compressive stress of 1.0 MPa
(rare load combination, also during construction stages) is expected from
In the following, because of lack of space, only two examples of continuous
Fig. 24- First balanced cantilever of Left Embankment viaduct, approaching cable-stayed bridge over Po River.
PC bridges are mentioned.
The first example is composed by the river embankments approach viaducts to the cable stayed bridge over Po River [2]: five spans on the left side
for a total length of 260 m (Fig. 24) and three spans on the right side for
a total length of 130 m.The decks are three cells box girders, built by balanced cantilevers from central piers: a couple of segments at one time is
casted in situ and post tensioning cables are tensioned.
Then the remaining gaps of 1.0 m long are concreted and post-tensioning
cables are laid in the spans to join together, two by two, the cantilevers. For
the Italian State Railways, this is the second example of this kind of structure and method of construction (the first one being built for the RomeFirenze line).
The second example refers to the Modena continuous bridges to overpass
two rivers, a highway and a railway Junction in the Modena System of
Viaducts, for a total number of nine single-track continuous beams.
All nine bridges have span lengths of 40-56-40 m, the same outer crosssection of the lower way U decks of Modena simply supported spans (type
f), with higher webs on central piers.They are built in three segments casted
in situ e prestressed with 40 mm bars, coupled at the end of each segment.
Two methods of construction were experimented for the same structural
solution, because of the different environmental conditions: for two of these
beams, the segments were casted on a formwork, scaffolding from the
ground with two temporary bars for each joint, in order to help the partial
structure during construction conditions.The other seven obtained the same
effect with a formwork hanging from steel box beams (over Panaro River)
or truss beams (over Secchia River and on Brennero highway, see Fig. 25)
and the end support of the beam applies nearly the same reaction of the
temporary bars.
This paper has described the strategic choices and the most relevant
structural and infrastructural features for the design and construction of
the new Italian high-speed railway network. Most topics concerning simply supported prestressed concrete decks for new high-speed lines have
Fig. 25- Continuous PC beam of Modena viaduct over Brennero highway, Milano-Bologna line.
CONSTRUCTION - Civil engineering Works
been analysed, while some information on continuous PC beam have
been introduced at the end.
Deck with precast beams and cast in situ slab is the most common choice due to its flexibility (type “b” is actually the most used structure for bridge decks) but full span pre-casting is the most important future trend.
The mentioned principles of design, the so-called multi-layer protection
system, the method of construction, which deeply influences design choices, and the tendency to experiment models and tests before every first
realization, have been recognised as strategic factors for good results in
design, construction and management of railway infrastructures.
Italian standard for railway bridges: Istruzione F.S. n. I/SC/PSOM/2298 del 2.6.1995 “Sovraccarichi per il calcolo dei ponti ferroviari - Istruzioni per la progettazione, l’esecuzione e il collaudo”,
Final review 1997.
Petrangeli, M. P.,Traini, G., Evangelista, L., Della Vedova, M.The cablestayed bridge over Po River: design and construction. Proc. of the
2nd International fib Congress, 5/8 June 2006, Napoli.
CONSTRUCTION - Civil engineering Works
Figures, deadlines and picture in §1.Introduction are drawn from
www.tav.it, revision April 2004.
Italian standard for railway bridges in seismic areas: Istruzione
FS 44/b “Istruzioni tecniche per manufatti sotto binario da vostruire in zona sismica”, Final Review 14/11/1996.
Italian standard for railway line maintenance: Istruzione FS 44/c
“Visite di controllo ai ponti, alle gallerie ed alle altre opere d’arte del
corpo stradale: frequenza modalità e relative verbalizzazioni”, Final
Review 16/02/1994.
HS railway cable-stayed bridge
over the Po river
Italian High-Speed Network
HS railway cable-stayed bridge over the Po river
Bologna Milano High Speed Line near Piacenza
T.A.V. S.p.A. Concessionaire for the design and
construction of the Italian HS system for RFI
Prof. Ing. Mario Paolo Petrangeli
Structural engineer
Prof. Ing. Mario Paolo Petrangeli
Prof. Ing. Mario Paolo Petrangeli
Management Contractor
CEPAV UNO- Syndicate formed by ENI per l’Alta
General Contractor
A.S.G. Scarl – Syndicate formed by Aquater,
Snamprogetti, Grandi Lavori Fincosit
Winner A.I.C.A.P. Award 2009 for Structural Concrete Works - Category “Civil Engineering”
The new railway linking Bologna to Milan is part of the High Speed Lines Italian network. It crosses the Po near Piacenza in a section where the river is usually about 350 m wide, up to 1 km
between the main embankments. The bridge will be 1200 m long, 400 m to cross the ordinary
riverbed, an obliquity of 22° resulting between the tracks and the river. Two approach viaducts,
respectively 6 and 4 km long, complete this work, the most important of the whole line.
Four main spans of 96 m were proposed in the preliminary design to satisfy the navigation requirement, and two solutions, both with prestressed concrete decks, were selected after a first study,
but the competent Authority for the Environment insisted in eliminating the central pier, a 192 m
main span resulting for that.
It is one of the longest prestressed concrete railway span in the world operated at a speed up to
300 km/h.
Three types of structure are present in the crossing in addition to the standard 14 km long approach viaducts placed outside the upper banks: the cable stayed bridge, 12 simply supported decks
on the right bank and two continuous p.c. box girders necessary to overpass the main embankments.
The decks are subdivided in such a way that two joints in the rails are necessary to keep the expansion length within the allowable limits. This is the only exception along the jointless HS Railway
Italian Network.
The relevant part of the crossing has a 192 m central span and two 104 m long side spans.
The deck is a p.c. continuous box girder with the fixed point at one tower, sliding bearings at the
second tower and at the transition piers. Expansion lengths of 296 and 104 m derived from this
arrangement of the bearings, joints in the rails so being required.The height of the cross section is
constant and equal to 4,5 m (L/42,7) along the central span; it varies and decreases to 3,70 m in
the side spans, in order to fit with the other decks.
The towers are 60 m high from the footing, 51 m from the deck. The top of the towers, where
the stays are anchored, is a steel-concrete composite structure.
The stays are made of 55 to 91 zinc-coated, singularly greased and sheathed 0,6” super strands.
The total amount of steel for the stays is 410 tons, corresponding to about 66 Kg per square meter
of deck.
The foundation of each tower has the footing (shaped to reduce the drag force) supported by
28 piles, 2m diameter and 65 m long.
Derailment of railway vehicles. Two accidental design situations have been considered with
respect to the stays:
- collapse of two consecutive stays along one side due to the derailment of a vehicle: the bridge
must remain in service with one design train over the track nearest to the injured side and one
passenger train ( 40 KN/m ) over the other, the thermal effects being excluded;
- the consecutive stays collapsed are three: only the effects due to one design train is taken into
Dynamic analysis.Three different trains (ETR 500,TGV, ICE) have been considered for the dynamic analysis considering the dynamic behaviour of the vehicle as well as the irregularities of the
Combined response of structure and track. The effects resulting from variable actions have
been taken into account according to prEN 1991-2 and for two limit stiffness of the foundations.
Seismic analysis have been carried out in the elastic range according to the Italian Railway
Specifications and EC 8. Because of the low seismicity, seismic actions did not influence the design
of the bridge but in a few sections in the upper part of the towers, while they were relevant for the
bearings and the joints.
All the decks have been built by cantilever method with cast in situ segments, but the 13 simply
supported spans.
CONSTRUCTION - Civil engineering Works
• 1- General view of the whole bridge.
A number of physical tests have been executed to assess the theoretical assumption, the most outstanding being: (i) test on a half scale
model reproducing a segment of the deck with a stay anchorage, carried on in the yard; (ii) fatigue test on a full scale model of the steel box
embedded in the upper part of the tower to anchor the stays and (iii)
fatigue tests on three stays (composed by 55, 73 and 91 0,6” strands)
complete of anchorage. Both fatigue tests were carried on in the EC
Joint Research Centre of ISPRA.
Due to the importance of the bridge, a large number of sensors have
been permanently placed on it. The monitored quantities are: loads on
the piles, stress and temperature in the most representative sections of
the deck and the towers as well, forces transmitted by a number of stays
and bearings, geometrical data like the angular rotation of towers and
the deflection of the decks and, finally, the scour near the piers in the
riverbed. Both sonar and magnetic devices have been installed to detect
scour. All the data will be collected inside the cable-stayed deck and from
there automatically transmitted to a remote office located in Bologna
that will manage the monitoring system of the bridges of the line.
CONSTRUCTION - Civil engineering Works
Italian High-Speed Network
CONSTRUCTION - Civil engineering Works
Completion of the towers and access viaducts. 3-4 Construction of the deck
•by 2-segments.
5- Pouring the crown segment. 6- Tower elevations.
CONSTRUCTION - Civil engineering Works
“Piacenza” viaduct
Italian High-Speed Network
Piacenza viaduct
Piacenza, Italy
Ferrovie dello Stato S.p.A.
T.A.V. S.p.A.
Higher supervision
Italferr S.p.A.
Dott. Ing. Villa
General Contractor
Cepav uno
In the Milan-Bologna stretch of the high-speed/high load railway line the Piacenza Viaduct runs
a total length of 5103 m. The structure breaks down into two stretches, Piacenza1 and
Piacenza2, their lengths 2522 m and 2581 m respectively.
The first stretch features a two-light caisson deck and a steel-concrete mixed-structure span 50
m long.
The second stretch, like the first, also comprises a two-light caisson deck plus two mixed-structure spans of lengths 38 and 40 meters.
The prestressed-concrete spans consist of monolithic precast two-light caissons with post-tensioned prestressing cables.
The caissons weigh 975 tons each and are 33 meters long.
The section depths/thicknesses are 45 cm for the webs, 37 cm for the deck slab and 30 cm
for the bottom slab with increased depth at 60 cm from the ends.
The prestressing cables are 24 post-tensioned cables, electrically insulated by the use of insulated head ends and HDPE sheaths. Of them, fifteen 19-strand cables are placed in the three
webs, with parabolic trajectories, while the other nine cables, 12-strand, are placed in the bottom slab on a straight-line trajectory. The strands used are class fptk 860 MPa, having a rated
section of 139 mm2.
Maximum pier height is 12 meters.
The foundations, deep, are built on piles.
The caissons were worked up in the plant, in the following phases: creation and placement of
the slack reinforcings and of the forms, a concrete pour carried out with the aid of truck mixers
and of two truck-mixer pumps positioned nearby the head ends, threading and tensioning of the
cables and, finally, hauling the caisson to the launch area. This last phase was carried out by a
special launch car weighing 900 tons.
CONSTRUCTION - Civil engineering Works
the launch caisson. 2- Hauling the caisson. 3- Launching the cais•son.1-4-Preparing
Mounting the side wings. 5- Viaduct cross section. 6- Construction phases: a)
pour of the bottom slab; b) pour of the septums; c) pour of the top slab; d) folding
back the internal forms for their removal. 7- Cross section through the two-light caisson. 8- Longitudinal section through the two-light caisson, showing prestressing cables.
CONSTRUCTION - Civil engineering Works
“Modena” system viaducts
Italian High-Speed Network
Modena viaducts
Modena, Italy
Ferrovie dello Stato S.p.A. (State Railway System Inc.)
T.A.V. S.p.A.
Higher Supervision
Italferr S.p.A.
Prof. G. Macchi (Final design), Ing. Sangalli
(Construction design)
General Contractor
Cepav uno
The Modena Viaducts system belongs to the high-speed/high-load Milan-Bologna line and has a
total length of 25,182 meters, of which 23,958 m built with precast segments and 1224 m
with in situ-poured continuous beams. The Table gives the names and the principal characteristics of the viaducts belonging to the “Modena” System.
All these viaducts have in common a type of deck known as the Omega, featuring a U beam
bearing the track on its flange for each track.This particular cross-section conformation has the
advantage of mitigating noise emissions by partially enclosing the train within the structure
through the bearing walls, which also act as noise barriers.
What is special about these viaducts is the particular precasting process used, which enabled
installation, during full production (between 2003 and 2004) of more than one segment per
day with a peak of 52 segments launched in July 2004.
Besides the 767 plant-precast (isostatic) segments, the structures call for an additional nine statically-indeterminate continuous-beam stretches of length 136 m, with spans of 40-56-40 m
(statically indeterminate), for a total of 27 spans. In the statically-indeterminate stretches too the
cross section is still the Omega type, with projections in the zones where the bending moment
is negative. The statically-indeterminate members are poured in forms hanging from overlying
The typical viaduct span in the isostatic stretches is a precast prestressed-concrete member
having an open cross profile.The precasting, for a single-track, is 9 m wide and 3.5 m deep, weighing a total 690 tons of which 33 tons are steel (slack reinforcing and prestressing). The outside surface of the deck features longitudinal channels and four r.c. support elements solidly joined to it, the internal surface is smooth and encloses the trackway.The decks were all built using
an Rck 45 MPa concrete.
The (circular section) piers are 3.5 m in diameter with heights from 5.45 m to 12.20 m.Their
sections display three typologies: hollow with a 50 cm wall-depth, hollow but filled, and full-section. In the statically-indeterminate stretches the central piers are 4.50 m in diameter with full
circular sections.The piers are built of Rck 35 MPa concrete.
The foundations were built in two ways: on 1500 mm diameter piles of lengths between 35 and
50 m, and on diaphragm septums, 120 cm thick and 36 m long.The footings have dimensions
of 9.5 m x 16.7 m x 2.2 m and were built of Rck 30 concrete.
CONSTRUCTION - Civil engineering Works
Longitudinal section through deck. 2- Cross section through isostatic portion.
through statically-indeterminate portions showing enlargements nearby
the areas of negative moment. 5- Precast segment for the isostatic span. 6- Launch of the isostatic segment. 7- Statically-indeterminate portion.
Isostatic beams, no.
Viaduct name
Total lenght (m)
Brennero - Odd track
Brennero - Even track
Modena Interconnection - Odd track
Modena Interconnection - Even track
Modena - Odd track
Modena - Even track
Panaro - Odd track
Panaro - Even track
CONSTRUCTION - Civil engineering Works
“Savena” Viaduct
Italian High-Speed Network
Savena Viaduct
Bologna – High-Speed Florence-Bologna Railway Line
T.A.V. S.p.A.
Studio Sintecna – Prof. Ing. Giuseppe Mancini
The Savena viaduct is the crossing of the river of the same name by the Florence-Bologna HighSpeed Railway Line on a single double-track deck.This structure’s main characteristic lies in the
deck’s structural depth (below the ballast), which is just 80 cm for a 63.5 m span. To achieve
this slenderness design opted for a through-arch in which the deck is the structural system’s tendon. The deck plate is thus so shaped as to create in its central zone the basin containing the
ballast, below which the structural depth is 80 cm. Laterally the depth was suitably increased
so as to correctly house the prestressing-cable heads.
The deck is suspended laterally and two alignments per side of hangers, connected to two r.c.
arches, inclined toward the interior of the bridge and connected in their turn by three r.c. crosspieces. In order to meet the design performance requisites, consisting in total prestressing in service, at the deck edges in the main directions two-way prestressing had to be applied to the
plate. Its effects were evaluated using a finite-elements mathematical model consisting of shell
elements. The hangers are connected to the deck through pre-tensioned prestressing bars and
anchored in the soffit.The bridge’s construction was carried out in the following phases:
• the three-phase deck pour on the embankment behind the abutment and the deck’s progressive crosswise prestressing on the ground; then the creation of the longitudinal prestressing;
• construction of two provisional piers in the river;
• a forestarling 22 m long is mounted;
• thrust of the deck up to the next abutment;
• the mounting of centerings and forms for the in situ pour of the arches;
• mounting and adjustment of the hangers, with the deck being detached from the provisional
• demolition of the provisional piers and their foundations.
CONSTRUCTION - Civil engineering Works
• 1-2- Deck plan, and cross section through viaduct.
CONSTRUCTION - Civil engineering Works
Italian High-Speed Network
3- Construction of the provisional riverbed piers. 4-5 Detail of the arch-plate join
and the reinforcing. 6- Installation of the arch reinforcings. 7- Forestarling for the
construction of the deck. 8- Anchorage of the hanger on the deck. 9- Bridge side
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
“Caivano” variation structures
Italian High-Speed Network
For the construction of the Rome-Naples High-Speed line through the municipality of Caivano,
the railway had to be inserted with the least possible environmental impact on the territory. This
led the customer, RFI-TAV, to work out “structures” that would be an addition to the territory.
The variation, which runs some six kilometers between stations 200 km and 206 km of the
Rome-Naples High-Speed Line substantially comprises three principal structures:
- the conical-capital viaduct;
- the multiple-arch viaduct;
- the two cut-and-cover tunnels Caivano 1 and 2.
Rome-Naples High-Speed Line
Caivano, Naples, Italy
R.F.I. S.p.A. (Italian Railway System Inc.) - T.A.V. S.p.A.
(High-Speed Train Inc.)
Prof. Remo Calzona CE
Società Italiana per Condotte d’Acqua S.p.A.
Year of completion
March 2006
Rome-Naples High-Speed Line
The conical-capital viaduct
The structure is 203 m long and serves to overpass three interfering lines having centerlines
strongly skew to the railway line: the Turano-Biferno aqueduct, the Secondigliano sewer line and
the SS87-bis highway. The positioning of the vertical elements follows directrices parallel to the
centerlines of the interferences, the deck is created by an orthoanisotropic plate, woven crosswise to these directrices so as to minimize the bearing structure’s span. Deck characteristics
and constraint scheme: two-span frame with an orthoanisotropic plate fixed-jointed to the
Soil conditions, water table: soils of volcanic origin featuring a first 8-10-meter stratum of fairly
well compacted pyroclastites, an underlying stratum six meters thick of tuff having lithoid characteristics and a successive stratum of more deeply densified pyroclastites.The water table lies
at an average four meters below site level.
Degree of seismicity: the area is classified as a second-category seismic zone, so that its degree
of seismicity, S, is 9.
Foundations typology: deep piles 1200 mm in diameter.
Construction procedures: the pier foundations and standing structure were in situ poured. For
the deck, resort was had to precasting long-span predalles; the precast beams were prestressed,
and an in situ completion pour was made for the slab.
Caivano, Naples, Italy
R.F.I. S.p.A.(Italian Railway System Inc.) - T.A.V. S.p.A.
(High-speed Train Inc.)
Prof. Remo Calzona CE
Società Italiana per Condotte d’Acqua S.p.A.
Year of completion
March 2006
Rome-Naples High-Speed Line
Caivano, Naples, Italy
R.F.I. S.p.A.(Italian Railway System Inc.) - T.A.V. S.p.A.
(High-speed Train Inc.)
Prof. Remo Calzona CE
Società Italiana per Condotte d’Acqua S.p.A.
Year of completion
The multiple-arch viaduct
The viaduct comprises a sequence of 74 parabolic-profile arches having a 33.00 m chord, arranged in two parallel files spaced 8.40 m apart. The arch height varies between 0.80 m at the
crown, including the depth of the slab, to 4.00 m at the springer. Its depth is instead a constant
1.00 m. The upper slab, 13.60 m wide, has a depth varying between 0.60 m at midspan to
0.50 m at the ends. Overall viaduct length is 2400 m.
Deck characteristics – constraint scheme: reinforced-concrete arch.
Soil characteristics, water table level: soils of volcanic origin, the first stratum of pyroclastites of
medium density varies in depth, being greater at the start of the viaduct and lessening towards
its end. Right below it is a tuff bench six meters deep and then the pyroclastites are taken up
again, more deeply densified.The water table lies at an average four meters below site level.
Degree of seismicity: the area is classified as a seismic zone of second category, so that the
degree of seismicity, S, is 9.
Foundation typology: deep piles an average of 35 m long and diameter 1500 mm.
Construction procedure: to build the standing structure a form was used created especially for
the purpose. It enables building in a first phase both arches and, after their curing, the slab in a
single pour.
The cut-and-cover tunnels Caivano 1 and 2
In order to underpass the two interfering roadways, motorway A1 and the Nola-Villa Literno
superhighway, two gigantic monoliths were designed, weighing 12,000 and 8,000 tons respecCONSTRUCTION - Civil engineering Works
tively, which were translated underneath
the two aforesaid interferences. Completing the job are the cut-and-cover rain tunnels, which at the mouths prevent the
entrance of rainwaters.
Soil characteristics and water table: soils of
volcanic origin. The first stratum of medium-dense pyroclastites varies between 67 meters of depth; the next 5 m down sees
a tuff stratum of good mechanical characteristics, and deeper down yet are again
dense pyroclastites. The water table lies at
four meters below site level.
Degree of seismicity: the area is classified
as a second-degree seismic zone, so that
its seismicity, S, is 9.
Construction procedures: the monoliths,
115 m and 70 m long, were poured in situ
in special forms and then translated using
a thrust system that enabled the translation of the greater of the two caissons in
just six days.
CONSTRUCTION - Civil engineering Works
Tunnels in the Florence-Bologna stretch of
High-Speed Line
Italian High-Speed Network
Tunnels in the Florence-Bologna stretch of
High-Speed Line
The design and construction of the tunnels were carried out under the A.DE.CO-RS (Analysis of
Controlled Strains in the Rocks and Soils) design methodology.
This approach was used to make up the base-contract project and then the construction design.
This tunnel design and construction methodology avoids the limitations of traditional approaches, by supplying the possibility of industrializing the tunnel advances even when traditional
driving methods are applied, in complex geomechanical contexts and with covers running from
a few meters to six hundred.The tunnels drive through sundry Apennine geological formations,
consisting mainly of marls, sandstones, scaly clays and limestones. In driving the running tunnels
the following methodologies were adopted:
• “Traditional” driving by the use of traditional machines (jumbos, excavators, jackhammers,
power shovels), except for the Ginori service tunnel, the only stretch in which wholly mechanized driving was carried out (by a shielded TBM). Overall some 17 million cubic meters of material were excavated.
• Full-section driving after consolidation of the core-face (where necessary), especially under difficult stress-strain conditions.
• Limitation of soil decompression by means of pre-confinement operations on the core-face or
confinement of the cavity (sub-horizontal jet-grouting, fibreglass plastic structural elements in the
core or around the cavity, shotcrete and steel ribs, radial bolts, etc.).The operations were defined
depending on the strain behaviour of the advancing core-face, evaluated during design phase,
described according to the following three behaviour categories: A= stable front, B= short-term
stable front, C= unstable front. Corresponding to each category is a tunnel section type for the
driving and advance.The percentage distribution of the tunnel section types applied is as follows:
type-A sections (35%), type-B sections (53%), type-C sections (12%).
• Final concrete and reinforced-concrete lining.The final linings (roof, sidewalls and inverted-arch)
were in situ-cast concrete of class C25/30 in depths of 60-100 cm. On the whole seven million
cubic meters of concrete were applied.
Duration of the construction phase: 13.5 years (from July 1996 to December 2009) with advance rate of one to five meters per day and up to 2000 meters per month on thirty faces at the
same time.
Florence-Bologna stretch
R.F.I. S.p.A. (Italian Railway System Inc.)
T.A.V. S.p.A. (High Speed Train Inc.)
Higher Surveillance
Italferr S.p.A.
MAIRE Engineering
Rocksoil S.p.A.
General Contractor
Year of completion
Main features
Total length of the High-Speed Florence-Bologna
railway stretch: 78.5 km
Underground works:
- 9 double-track driven tunnels having multi-centered
sections with a driven area of 140 m2 and a net
area of 82 m2, their lengths running from 693 m to
18.2 km;
- 13 access windows for a total length of 9.3 km;
- one service tunnel 10.65 km long;
- 2 single-track interconnection tunnels (total length
2.2 km);
- 3 cut-and-cover double-track tunnels 2.8 km long;
- one large underground chamber for switching
space, one underground interconnection chamber.
CONSTRUCTION - Civil engineering Works
1- Pianoro Nord area: Savena bridge and the portal of the Pianoro Tunnel. 2•Northern
portal of the Monte Bibele tunnel. 3- Firenzuola chamber. 4- From the
northern Morticine portal to the southern Borgo Rinzelli portal. 5- Southern portal
of the Sadurano tunnel. 6- Southern portal of the Raticosa tunnel.
CONSTRUCTION - Civil engineering Works
New stations for Italian High-Speed
Italian High-Speed Network
In all cities the construction of a new railway station is a special event, if only in that unlike an airport or other man-made transportation facilities, it
constitutes a “port within the heart” of the city itself.
In Italy station construction has marked the stages in the modernization of the country; the stations of the main Italian cities were the starting point
of the “journey” and as such were often constructed as authentic monuments.
The new stations that are emerging due to the advent of high-speed rail transportation are being located in the cities according to different criteria
than in the past, often being designed to reunite parts of the city separated by the railway lines and to contribute to their redevelopment and enhancement. Some of the world’s leading contemporary architects have been commissioned to design stations that are full of life, light and sound -in a
word, stations with a definite soul, and one that they are able to project: no longer an anonymous “non-place”, a desolate point of transit and metaphor
for lone travel, the new stations are “the place”, and, similar to how a town square encourages aggregation, dialogue and leisure, they embody the
very concept of urban community life. Railway stations will be home to bookstores, cafes and shops and become venues for music, art and culture
so that they come to be looked on by the public as places in which social relations are played out.
On the horizon are the new HS stations of Naples Afragola designed by Zaha Hadid, Rome Tiburtina designed by ABDR-Paolo Desideri; Florence
Belfiore designed by Norman Foster; Bologna designed by Ricardo Bofill; Reggio Emilia designed by Santiago Calatrava and Turin Porta Susa designed
by Arep Group. Meanwhile, plans and works are in progress for the redevelopment of the historic mainline stations to adapt them to new management and operational requirements and enhance the cultural heritage they represent.
CONSTRUCTION - Civil engineering Works
HS Bologna Station – the map. 2- Structural sketch of the new Bo•logna1- TheHS new
Station (drawing by P. Bellotti). 3- The transversal section. 4- The new HS
Bologna Station work site.
Bologna High-Speed train station
R.F.I. S.p.A. (Italian Railway System Inc.)
Design and structural engineer
Management Contractor
Italferr S.p.A.
Ricardo Bofill
Astaldi S.p.A.
The Bologna HS Station extends under a section of the track array of the existing central
station into a large underground chamber
approximately 640 m long, 42 m wide and 23
m deep. The new underground station will be
split into three levels.The HS tracks will be laid
on the bottom level, passenger services and
commercial activities will be located on the
middle level, while the upper level will be reserved for vehicular traffic to the
station and as parking space.The middle level will also be used as an exhibition area for the numerous archaeological finds (including a road, a furnace and burial chambers) found during excavation. The HS Station was
designed by the Catalan architect Ricardo Bofill who adopted innovative
solutions of high architectural value. The supporting structures of the excavation, for example, are designed as “facing arches” that transfer earth
thrust to “spurs” placed at 12 m intervals: this results in large bulkhead sections free of structural obstacles both horizontally (12 m of clear space) and
vertically (24 m of clear space). This innovative solution exploits the facing
arches as architectural elements to project daylight onto the tracks creating
a vertical illumination effect similar to that found in Gothic architecture and
also allows the insertion of vertical communicating elements between levels.
An international design competition for the completion of the central station
and the redevelopment of the urban area in which it is located has been
recently awarded to the Japanese architect Isozaki.
Main features
Plan dimensions: 640m x 42m
Overall height: 23m (underground)
Pile perimetral bulkheads: 20,000 m2
Bulkheads, rider arches and spurs = 37,000 m2
Foundation concrete: 57,000 m3
Foundation reinforcing steel: 10,000 t
Concrete in standing structure: 100,000 m3
Reinforcing steel in standing structure: 15,000 t
Structural steel: 31,000 t
Floor-structure area: 80,000 m2
CONSTRUCTION - Civil engineering Works
•tion.1- Overall view. 2- Detail the steel-glass canopy. 3-4- Internal view. 5- Cross sec-
Italian High-Speed Network
Main features
Plan dimensions: 450m x 52m
Overall height: 26.5m
Bulkheads: 40,000 m2
Foundation concrete: 95,600 m3
Reinforcing steel in foundation: 22,800 t
Standing-structure concrete: 75,000 m3
Standing-structure reinforcing steel:
23,700 t
Structural steel (tons): 8,000 t
Floor-structure area: 35,000 m2
Florence High-Speed railway station
Former Macelli/Belfiore area of Florence, Italy
R.F.I. S.p.A. (Italian Railway System Inc.)
ATI Foster and Arup
Structural engineer
Definitive design: Arup
Final design: Nodavia
Foster & Partners
Management Contractor
Italferr S.p.A.
General Contractor
Nodavia, Ati Coopsette-Ergon engineer
The new Florence HS station will be built in a
nineteenth-century district of the city, that of
the former Macelli (slaughterhouses) area, and
is part of a wider-ranging project for the recuperation of abandoned industrial sites in various parts of the city.
The new station will not replace but form an
interchange node with the existing Santa Maria Novella station. The two stations will be
linked by a new tramway that will extend to
Peretola airport and by shuttle trains between
the Circondaria regional train station and
Santa Maria Novella station.
Like the great stations of the past, the new HS
station will feature large, mainly glass-covered
volumes. While construction technology and
trains have radically changed since the nineteenth century, large spatial dimensions and
quality remain indispensable requirements for
modern stations.The project was developed by
the architect Norman Foster.The station will be
built on two underground levels in a vast
chamber 454 m long and 52 m wide, and the
rails of the high-speed line will be located at a
depth of 25 m from ground level.
From a structural point of view the station is
characterised by 1.6 m thick vertical support
panel bulkheads reinforced by perpendicular
spine walls and strutted in such a way as to
allow ample clearance for the passage of light.
CONSTRUCTION - Civil engineering Works
• 1-2- Overall view. 3- Detail.
The new Rome Tiburtina High-Speed train
station – 1st functioning job segment
Rome, Italy
R.F.I. S.p.A. (Italian Railway System Inc.)
ABDR Architetti Associati – Paolo Desideri arch.
(Chief of job coordination)
Structural Engineer
Ezio Maria Gruttadauria CE
Studio Tecnico ABDR architetti associati
Contract Management
Italferr S.p.A.
ATI Coopsette / MECoop
Main features
Overall height: 36 m of which 26 m above
Bulkheads: 29,150 m2
Piles: 62,190 m
Foundation concrete: 42,450 m3
Reinforcing steel in foundation: 14,425 t
Standing structure concrete: 24,675 m3
Standing structure reinforcing steel: 7,415 t
Structural steel: 11,750 t of which 1,475 t for the
roof space-lattice structure
Floor-structure area: 54,000 m2
CONSTRUCTION - Civil engineering Works
1- General view. 2- Transversal section of the first part of the station. 3-4-5-6-7•8- The
work site.
Italian High-Speed Network
The underground station
The station was built in two phases under
two integrated projects, developed in different
periods. The first project is tied to the construction of the Turin through line for putting
underground the tracks running through the
city and for quadrupling them; this to be done
with the construction of the Porta Susa station underground. This station, essentially an
underground chamber conceived as an underground station, is in an advanced stage of
construction and is partially operating since
In what follows the principal characteristics of
both projects are set forth.
The Turin Porta Susa High-Speed train station
Turin Porta Susa, Italy
R.F.I. S.p.A. (Italian Railway System Inc.)
ITALFERR S.p.A. (Definitive), Ing. Campa (Final)
Structural engineer
ITALFERR S.p.A. (Definitive), Ing. Campa (Final)
ITALFERR S.p.A. (Definitive), Ing. Campa (Final)
Management Contractor
A.T.I. – Astaldi S.p.A. (Mandatary Group Leader) –
Vianini Lavori S.p.A. (assignor) - Impresa di
Costruzioni Rosso geom. Francesco & Figli S.p.A.
(assignor) – Di Vincenzo Dino & C. S.p.A.
(assignor).– Turner & Townsend Group Limited
Main features
Plan dimensions: 465 m x 45 m
Overall height: 12 m
Bulkheads: 2,500 m2
Foundation concrete: 20,000 m3
Reinforcing steel in foundation: 2,4 kt
Concrete in standing structure: 23,500 t
Reinforcing steel in standing structure: 3,4 kt
CONSTRUCTION - Civil engineering Works
• 9- The new station. 10- The structure completed. 11- Detail. 12- Overall view.
The new station building
The second project for completion foresee a
station building integrated into the new city
plan. The station also acts as an exchange
node with the underground MRT system.
The Turin Porta Susa High-Speed railway station
Turin Porta Susa, Italy
R.F.I. S.p.A. (Italian Railway System Inc.)
Design/Structural engineer/Architects
Joint venture of designers: AREP J.M. Duthilleul,
E. Tricaud (Group leader), arch. Silvio D’Ascia,
prof. arch. Agostino Magnaghi
Contract Management
R.F.I. S.p.A.
Joint venture of companies: Guerrino Pivato
S.p.A. (Group leader), BIT S.p.A.
Main features
Plan dimensions: 385 m x 33,60 m
Overall height:
• portion below street level: 10.19 m
• portion above street level: varying from 3 to
12 m
Floors above ground: 1
Floors below ground: 3
Gross pavement area: 39,800 m2
Underground parking area: 8,700 m2
CONSTRUCTION - Civil engineering Works
Italian High-Speed Network
Reggio Emilia High-Speed railway station
Reggio Emilia, Italy
The Emilia-Romagna Region and T.A.V. S.p.A.
Santiago Calatrava
Structural Engineer
Impresa Cimolai S.p.A. is making use of the
following designers: SETECO (steel) – SGAI
Impresa Cimolai S.p.A. – RPA (architecture) with
the supervision of architect Calatrava for the
artistic aspect
Contract Management
Italferr S.p.A.
Impresa Cimolai S.p.A.
Main features
Plan area: 483 m x 50 m
Overall height: 20 m average, since the roof has a
sine-wave profile and h varies.
Foundation concrete: 16,734 m3
Foundation reinforcing steel: 1,700 t
Standing structure concrete: 2,250 m3
Standing structure reinforcing steel: 430 t
Structural steel: 9,140 t (portals, arches and
Floor structure area: 7,728 m2 (loading platforms)
and 200 m2 (ground-floor business stalls)
CONSTRUCTION - Civil engineering Works
Naples – Afragola High-speed railway station
Afragola district - Naples, Italy
R.F.I. S.p.A. (Italian Railway System Inc.)
Definitive design - Zaha Hadid arch.
Final Design: SAIR – GEIE (European
Architecture, Town-planning and Engineering
Structural Engineer
Definitive design - Zaha Hadid arch.
Final design - SAIR – GEIE (European
Architecture, Town-planning and Engineering
Zaha Hadid arch.
Contract Management
Italferr S.p.A.
Main features
Plan dimensions (project area, including parking):
450 m x 350 m
Overall height: 26.45 m
Bulkheads: not present
Foundation concrete: 38,000 m3
Foundation reinforcing steel: 4,000 t
Standing structure concrete: 45,000 m3
Standing structure reinforcing steel: 5,200 t
Structural steel: 4,200 t
Floor structure area: 23,000 m2
CONSTRUCTION - Civil engineering Works
“Colletta” cable-stayed bridge
The Colletta cable-stayed bridge
Sangone creek, Giaveno (Turin), Italy
The District of Turin
A.T.I. [GEODATA SpA, Mario Petrangeli & Associati
Srl (Prof. ing. Mario Paolo Petrangeli, ing. L. Fieno,
ing. L. Pinchiaroglio)]
General contractor
A.T.I. (SISEA, COGEIS, FIP Industriale)
Stays (supply and work)
Mario Petrangeli & Associati Srl
After the floods of October 2000 that caused enormous damage, the province of Turin
appropriated funds for setting the riverbeds to rights and for reconstructing damaged
infrastructures, among which the bridge over Sangone creek. The impossibility of changing
the existing highway route, owing to the numerous constructions lined along it, compelled
design to conserve the old bridge’s position in both line and grade, and this was a heavy
constraint on the design of the new one. In making up the design, the alternatives evaluated had to eliminate the intermediate supports (their presence would mean the construction of deep-lying foundation structures) and increase bridge length by twenty metres, while
meeting the hydraulics clearances required and providing a wide range of choices of construction methods and an architectural value suited to the site’s environmental value. At the
end the choice fell on a cable-stayed bridge eighty meters long. The deck cross section, a
total 15.10 m wide, is the classical one for small-span stayed bridges: there are two main
girders, of rectangular section 2.0 m x 1.20 m, in situ-poured and lightly prestressed, on
whose centerlines are anchored the stays, with their heads placed in special places.The two
longitudinal beams bear the secondary crosswise weave, created with adherent-strand prestressed precastings placed at 2.50 m spacings.The pylon, a total 36 m high from its springer, consists of two blade uprights inclined by ten degrees to the vertical and connected by
an upper prestressed crosspiece, placed immediately below the stay anchorages. The pylon
bears eight pairs of stays that sustain the eighty-meter span and is itself anchored to the
ground by another four pairs of stays that are anchored on the mooring block. The pylon
base is a caisson foundation structure, such as to directly transfer the stresses below the
creek bed directly onto the bedrock.
CONSTRUCTION - Civil engineering Works
in which is seen too a bridge built for provisional traffic flows. 2- The brid•ge,1-frontPlan,view.
CONSTRUCTION - Civil engineering Works
Pour of the deck slab. 4- The completed pylon structure, awaiting removal of
5- The stays system in its end configuration: in view are the two
mooring blocks anchoring the stays, since the single-span bridge scheme is dissymmetric. 6- Cross sections through the deck: typical (left) and nearby the pylon
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Rio S’Adde viaduct
Road viaduct
Rio S’Adde – SS 129, Macomer, Sardegna, Italy
ANAS spa
Agency letting contractor
Province of Nuoro
General design
Ing. Gian Paolo Gamberini
Structural design
Ing. Pietro Paolo Mossone
Chief of supervision of construction
Ing. Gian Paolo Gamberini
General contractor
CONSCOOP, Forlì – Coop. Edile “Edile di Orgosolo
s.c.r.l.” (awarded contractor)
The structure, enabled the elimination of several kilometres of tortuous road that made connections
with the city of Macomer, in the heart of Sardinia, inconvenient. This was effected by a viaduct that crosses a valley 180 m broad. The nature of the place, imposed on design choices
aimed at maximizing the oeuvre’s harmony with its context.The structure comprises a singlelight box deck of variable section set on two slender piers, each composed of two septums
set close together. The deck breaks down into three spans of lengths 45, 90 and 45 metres.
The deck depth runsbetween 2.30 and 4.50 metres, while the depth of the box soffit slab
varies between 0.22 and 1.00 m.The thickness of the side walls is 0.50 m and the extrados
slab is 0.22 m deep. The crosswise width is 13.00 m at the upper slab and 6.50 m at the
lower. The piers each comprise two septums having a section of dimensions 1.00 m x 7.50
m and heights of 21.70 m and 25 m respectively.
The two septums are connected together at the base by crosswise septums 0.50 m thick.
Deck construction proceeded by symmetric balanced advance cantilevered from the piers
with pours of successive segments; these were then prestressed in pairs by the application of
post-tensioned cables. The deck was built of 43 segments. The two approach spans comprise ten segments each, and the centre span comprises two facing halves, each of which breaks
down into ten segments; there are also the crown segment and the two pier-head segments.
Each segment is 4.20 m long except for the centre span crown segment, which is 1.00 m
long.The deck is prestressed with post-tensioned cables.The pierhead segments were poured
on centerings, to create the base platform for the launch cars. From these segments construction went ahead symmetrically with two isostatic tees being built, joined by the crown
CONSTRUCTION - Civil engineering Works
Deck cross section at the pier. 2- Cross section through deck at midspan. 3•The1-procedure
for constructing the deck cantilevered from the pier. 4- The septums
composing the pier structure.
CONSTRUCTION - Civil engineering Works
• 5- Bridge grade profile.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Bridge over Vajont creek
Road bridge
San Martino, Erto (Pordenone), Italy
Erto and Casso Municipality
Structural design
Ing. Paolo Giovenale, ing. S. Rossi, Roma
Geothecnical works design
Prof. F. Colleselli, Padova
Chief of supervision of construction
Ing. P. Sommavilla, Belluno
Safety coordinator
Ing. A. Tenani, Belluno
Impresa costruttrice
Monti S.p.A., Auronzo di Cadore (BL)
The bridge over Vajont creek, which stands five kilometres in an air line from the dam, was
literally swept away by the wave, running uphill, generated by the slide of 260 million cubic
meters of material that detached from Mount Toc on October 9th 1963. Some three years
ago the Erto and Casso city administrations decided to start up reconstruction work of the
bridge and to improve the existing city road on the left bank of the creek.The solution approved was a prestressed-concrete box-girder bridge built cantileverwise by successive segments
starting from the right bank of the creek. In its definitive configuration the bridge is a total
87.25 m long and 7.50 m wide.
The net span of the bridge between its bearings is 74.50 m.The rest of the deck corresponds
to a stretch built on the right bank, that acts as counterweight, and anchorage for the stay
cables.The route in plan is in tangent, and its grade profile lies at 731 m above sea level.The
roadway section has a 4.50 m wide carriageway and a bicycle and foot path 1.50 m wide.
The deck consists of a prestressed-concrete single-light box whose depth varies from 6.94 m
at the bearings on the right bank, to 3.25 m in midspan. The width of the box is a constant
3.00 m. The upper slab is 7.00 m wide. The depth of the extrados slab is a constant 0.25
m in the center, between the (vertical) webs. It descends to 0.20 m in the cantilevered parts.
The soffit slab depth varies from 0.25 m in span to 0.70 m at the starting section of the first
segment. The webs are 0.30 m thick. The statics scheme of the structural system as a whole
is thus a fixed-jointed or simply-supported beam. The deck comprises a total of fourteen in
situ-poured segments.The abutment segment, 14.00 m long and weighing 930 tons, was poured in a specially built form bearing on the ground. The last segment, 2.75 m long and weighing 55 tons, was poured directly on a form mounted on the left-bank abutment.
CONSTRUCTION - Civil engineering Works
1- Longitudinal section through the bridge. 2-3 Phases in the cantilevered con•struction
of the deck by successive segments. 4- The last segment connecting with
the tunnel.
CONSTRUCTION - Civil engineering Works
• 5- The bridge in plan. 6-7 Testing and load testing.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
“Cesare Cantù” cable-stayed bridge
Cable-stayed bridge, crossing the Adda river
Between Calolziocorte (LC) and Olginate (LC), Italy
Province of Lecco
Eng. Angelo Valsecchi (Department of Transportation,
Province of Lecco)
Structural engineering
Prof. Carmelo Gentile, PhD, Eng. Roberto Gentile
Management Contractor
Department of Transportation, Province of Lecco
General contractor
Vitali S.p.A., Cisano Bergamasco (BG)
Year of completion
The “Cesare Cantù” cable-stayed bridge is a roadway bridge, crossing the Adda river between
the municipalities of Olginate and Calolziocorte (about 60 km north-east of Milan).The bridge,
opened to traffic on February 13th 2009, is the seventh crossing of Adda river in the Province
of Lecco and represents an important commercial link between the opposite banks of the
river in a very intensely industrialized area; the conceptual and executive design of the new
bridge were completed in 1999 and 2003, respectively.
The bridge consists of two H-shaped concrete towers, double-plane cables and a pre-stressed
concrete girder deck; the bridge girder is formed by a central span of 110.0 m and two lateral spans of 55.0 m, for a total length of 220.0 m.
The deck, made in C40/50 pre-stressed concrete, was cast in place and post-tensioned. The
deck is 11.50 m wide and consists of 2 two-cell box girders, 1.50 m high, connected by a
central slab and by a series of 24 transverse cross-beams, providing the lower anchorage of
the stay cables. The equally spaced cross-beams are 8.00 m apart and their width exceeds
of 1.25 m per side the width of the deck. The deck was designed to have a depth of 1.5 m,
corresponding to about 1/150 of the total length, so that a good transparency of the bridge
girder was attained from aesthetic stand point.
The deck is suspended from 48 stay cables, arranged in two planes in a semi-fan, held at the
top of the two main reinforced concrete portal towers in special welded steel tower-head
assemblies.The cast-in-place concrete towers (C32/40) are about 38.0 m high and consist of
two concrete piers, a lower concrete wall connecting the piers and supporting the deck, upper
steel devices providing the anchorage for the stay cables and a transverse steel truss connecting the upper part of the piers.
The bridge stays use 19 and 31 16 mm diameter strands of steel cable; each strand comprises seven 5 mm wires and is protected by a high-density polyethylene sheath.
The construction of the deck was carried out by using an unusual technique. First, a series of
large diameter steel pipes were placed on the bed of Adda river in order to allow the water
flow corresponding to the maximum expected flood level. Subsequently a stabilized embankment was constructed over the steel pipes and the river bed, so that the embankment was
used to continuously support the formworks for the casting of the concrete deck. It should be
noticed that the above construction technique, usually adopted for very small spans, turned
out to be cost effective; in addition, the stability of the embankment and the effectiveness of
the water pipes were successfully checked by the severe flood occurred in Italy during Spring
It is further noticed that, according with the seismic classification introduced in 2003, the seismic hazard at the bridge site was described by the ground design acceleration ag=0.05g and
the importance factor was assumed as gI=1.3. The seismic design of the bridge was carried
out in order to ensure: (a) elastic behaviour in the vertical and transverse directions; (b) nonlinear dissipative behaviour (by using base isolation hysteretic devices) in the longitudinal
It is worth mentioning that the reception tests of bridge included not only the procedures that
are mandatory in the Italian Code (i.e. extensive characterization of the materials and severe load tests) but, according to the international practice, also dynamic tests (involving deck,
towers and stay cables) during the construction phases and before the bridge opening.
CONSTRUCTION - Civil engineering Works
The cable-stayed bridge on the opening day (February 13 , 2009). 2- Eleva•tion,1-plan
and deck cross-sections of the bridge (dimensions in m).
CONSTRUCTION - Civil engineering Works
3-4-5- Construction phases of the bridge. 6- Aerial view of the “Cesare Cantù”
bridge. 7- View of the bridge central span during the reception load tests. 8- Details of the transverse cross-beams providing the lower anchorage of the
stay cables.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Bridge between La Maddalena and Caprera
La Maddalena
Bridge between La Maddalena e Caprera Islands
La Maddalena Island (Italy)
Italian Government
Prof. Ing. Gian Michele Calvi
Structural detailing
Prof. Ing. Gian Michele Calvi, Lombardi Reico srl (Ing.
Giorgio Pedrazzi, Ing. Carlo Beltrami), Ing. Matteo
Prof. Ing. Gian Michele Calvi
Contract Management
Ing. Valter Frascaroli, Ing. Matteo Moratti
General contractor
A.T.I. Dott. Mario Ticca S.r.l. – Sassari, S.C.I.R. S.p.A. –
Cagliari, Novaco S.r.l. – Sassari
Year of completion
The bridge replaces a temporary Bailey bridge connecting the islands of La Maddalena and
Caprera.The design bridge length is 52 meters, with three spans symmetrically arranged (two
lateral spans 13.5 m long and a central one of 25 meters). The rise of the arch is about 5
m. The bridge is composed by double cantilever truss elements, conceived to conceptually
allow the rotation of each side on its foundation to open the central span.The variable thickness of the arches that form each cantilever element is such that the vertical dead loads reaction is centred in each foundation. The steel truss elements connecting the upper and lower
arches have circular hollow sections (D=150 mm).The lower central span arch and the upper
deck merge at the centre in a flexural hinge transmitting shear forces. Axial forces and bending moments reactions are provided at each one of the central arch foundations, by three
bearings with tensile reaction capacity.Tension and compression axial force restraints are provided at each abutment. The use of high performance concrete allowed to reduce the structural thickness and to use post-tensioning tendons in the upper deck, to equilibrate part of
the tensile forces developed in the upper chord of the spatial trusses. The whole bridge has
been protected with white polymeric resin, to provide durability and at the same time to
improve the aesthetics of the bridge.
All design choices are essentially governed by environmental constraints:
1. the geometry of the fixed points derived from the historical heritage, as well as the idea of
an arch bridge and of a conceptually rotating structure;
2. the central span height was the best compromise between deck slope and boat clearance
space (5 m (h) by 8 m (w));
3. the arch shape and the reduced element thickness minimizes the environmental impact;
4. the horizontal reactions at the abutments and foundation had to be limited because of the
soil characteristics;
5. an improved structural durability was dictated by the marine environment;
6. the on-site work duration had to be kept at a minimum to minimize the interference with
the local traffic;
7. two longitudinal concrete ribs (20 x 50(h) cm) have the double purpose of protecting the
pedestrian traffic without inserting heavy guard rail elements, and of contributing to the upper
chord structural capacity;
8. the external light parapet complete the sailing boat reminiscence of the whole structure,
with only white resin and stainless steel used as finishing materials.
CONSTRUCTION - Civil engineering Works
1- Historical bridge in the same location (destroyed). 2- Details of the parapets
•of the
pedestrian walkway.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
3- Construction of the piles cap. 4- Detail of the active-anchorage zone of the
tendons. 5- Temporary support and scaffolding of the central arch. 6Joint at midspan: fixed ends of the tendons on both sides.
CONSTRUCTION - Civil engineering Works
“Don Bosco” bridge at Arenaccia.
Architecture, white as light
Don Bosco bridge at Arenaccia
Naples, Italy
City of Naples – RUP Giuseppe Pulli arch.
Design and yard management
Prof. Antonio De Luca CE
Giuseppe Mautone CE, Alfredo Sasso CE,
geom. Corrado Esposito
Architectural consulting
Prof. Fernanda De Maio arch., Gianluca Marangi arch.
Artistic consulting
Mariangela Levita
Structures consulting
Prof. Attilio De Martino CE
Geotechics and foundations
Prof. Carlo Viggiani CE
General proof testing
Prof. Roberto Ramasco CE
General contractor
Joint venture of 3 parties: Fico Costruzioni srl,
Amato Trivellazioni srl, Fico Giuseppe
Concrete supplier
IMECAL s.r.l.
Mosaic supplier
Year of completion
The Via Arenaccia is one of the Naples street system’s main arteries, a role it has maintained
as the centuries passed.
In 1999 the City of Naples decided to replace the old bridge with a new structure that would
re-establish, in its entirety, the city’s old connection – as borne witness to by 18th century maps
– with its northern districts.
The project is highly complex and has many critical features, arising from the bridge’s position in relation to other existing buildings, in an old urban area that is today densely populated. A few of its interactions with existing infrastructures may be mentioned: two 1000 mm
pipes under pressure, two sewer trunks six meters wide, 24,000 telephone cables, many electrical cables of medium and high voltage, and gas lines.
The very slender deck (its depth at midspan is 50 cm) has the shape of a strongly depressed skew vault.The net span between the bridge abutments, measured on the skew, is 26.90
m, with a width of 19,85 m. The slenderness coefficient, L2/f, is 1450, a very high value for a
concrete arch, putting it at the top of the list of constructions of this type internationally.
The heavy thrust exerted by the arch deck is taken by the very sturdy caisson-form abutments, which constitute two control rooms needed to create the hydraulic bypass.
Special care was taken with durability, by choosing the best material that could be used for
the purpose: self-compacting concrete, and by seeing to its physical and mechanical protection through the use of a cladding, long-lasting and easily maintained.
The concrete protection was thus converted into a design and architectural point of departure to best fit the infrastructure into its urban context. The choice was to wrap all surfaces:
piers, abutments, vault, shelf girders, etc. with glass-mosaic tesseras forming a design, its purpose also to give the structure uniformity and continuity.
Bridges built in urban areas too often feature elements lacking order, disconnected and
various: beams, crosspieces, bolts, welds, aprons, sails, little shelves, shelf girders, railings, guardrails, meshes. These elements often form a disorderly universe that allows no possibility of
cleanness of form in creating an insertion into the urban fabric. The infrastructure thus becomes, too often, synonymous with poor architectural quality. Concrete, with its vocation for
the continuous structure, lends itself to eliminating the diversity of elements: piers, main and
secondary girders, crosspieces, shearbracings etc. The mosaic thus offered the possibility of
creating a uniform whole.
The decision to clad the bridge with mosaic tesseras gave the new infrastructure complex a
very obvious character, making it readable from the road crossing it even at long distances.
Design had, in this case, set up, after consulting the painter Mariangela Levita, an abstract
design of great size in tones of black, grey and golden-yellow on a white background, the idea
to transform the passageway into a true promenade amid art and architecture.The design
on the down-slope abutment bends then towards the new stairway both to mark the thickness of the abutments (six meters deep) and to suggest the continuation of the walk towards
the city’s upper level.
The protection of the concrete surfaces, with a view to the durability and sustainability of a
public work, formed the point of departure for an artistic operation and for the best fit of the
infrastructure into its urban context, where the bridge becomes something that does not arrogantly state its presence but that rather seeks to fit its various parts in, discreetly.The mosaic
is like a mantle in Christo’s installations, which enwraps all parts of the bridge to give it conformity and continuity. The whole (mosaic, railings and light poles) as white as light.
Project co-financed by the European Union FERS European Fund for Regional Development
P.O.R. Campania 2000-2006 Measure no. 5.1
CONSTRUCTION - Civil engineering Works
project design. 2- Cladding of the upslope abutment. 3- View of the down•slope1- Theabutment.
CONSTRUCTION - Civil engineering Works
of the bridge with the stairway connecting Via Arenaccia and Via Don Bo•sco.4-5-6ViewBridge
longitudinal section and plan.
CONSTRUCTION - Civil engineering Works
Photo: Luciano Romano
CONSTRUCTION - Civil engineering Works
Viaduct for State Road (SS)23
Road viaduct
Pinerolo – traffic circle grafting SS23 to SS589
Agenzia Olimpica Torino 2006
Studio SINTECNA – Prof. Giuseppe Mancini CE
Torino Scavi Manzone S.p.A. – General construction
Year of completion
The viaduct is part of the upgrading operation on the regional highway Laghi di Avigliana, the former State Road 589.This secondary extraurban artery (class C in the CNR standards) is being converted into a principal extraurban highway (class B), in the stretch in the City of Pinerolo from station 27+800 km to station 30+250 km, by improving its exchange systems with the main communication ways (principally State Road no. 23). At the intersection between SS 589 and SS 23 the
construction of a traffic circle was envisaged, having three lanes and an internal radius of 91.00 m,
to distribute the traffic into the various directions. Called for too is the construction of an overpass,
SS 23 flying over the new traffic circle, thus freeing the bypass traffic from all the other flows. At the
crossing with Corso Torino there was a grade intersection. Since the object was to free the state
highway's route of the urban fabric, the construction of a high-speed road in cut was proposed, so
as to underpass the crossing.This will be placed above site level and will be a two-lane traffic circle
having an internal radius of 31.00 m. It will comprise two simply-supported viaducts in curve. Both
viaducts display a section involving a ribbed prestressed-concrete plate of maximum depth 1.30 m.
Its outside profile is rounded to make it fit better into the landscape.This solution, which today receives everybody's thorough approval for its safety and durability, offers the following advantages:
• it has a significant structural mass that, even with reduced spans, makes it insensitive to the dynamic effects induced by the transit of vehicles;
• it can harmoniously take on any form in plan;
• it has a high guarantee of durability, because of its massive structure devoid of thin elements, because of the reduced extent of its exposed surface as compared with more traditional solutions (beams
plus slabs), and because of the adoption of HDPE sheaths for the prestressing cables, wholly impenetrable by the chlorides of antifreeze salts, which will certainly be used for a considerable period of
the year;
• it provides the possibility of forming the section at design’s pleasure, and thus of facilitating its fit
into the surrounding environment. Proposed for the case in question is a crosswise profile having
sizeable curved fairings and a central zone of lower width, which gives the impression of high slendemess to an observer standing on the ground around it, even if his point of observation is but a
short distance from the structures;
• there is the possibility of pursuing the structural shape with the finish works (parapets, barriers,
protections for the underlying crossings), so shaped as to harmonize the structure-finishings complex.
The SS 23 viaduct on the traffic circle grafting it onto SS 589 has six spans.The two side spans at
the end are 28 m long, and the four central spans are 33 m long.The structure is continuous-beam
with intermediate bearings. Considering its height (not far above ground) and the reduced interference with the road system during construction phase, its construction was envisaged as in “standard
shoring”, that is with formwork and strutting bearing on the ground, through a slab to distribute the
forces.The construction methods call for a complex centering involving six successive phases, each
of which involving the tensioning of the prestressing cables envisaged for the construction phases.
Finally, when the pours have been completed, a further prestressing will be introduced into the final
statics scheme through the use of a second series of cables, having two different types of trajectories.The piers have a structural geometry with a profile faired into the deck’s and a very limited longitudinal dimension (one meter) so as to not form a barrier effect for the observer glancing’at the
panorama from below. On the outside of the viaduct a service sidewalk is envisaged with a parapet
whose function is also that of noise barrier. It is composed of metal uprights and coloured polycarbonate panelling having a curved profile so as to fair into the deck section. In order to facilitate
inspection and maintenance of the works inspection spaces are provided for the prestressing head
ends.There is also the possibility of replacing the deck bearings without limiting traffic.
CONSTRUCTION - Civil engineering Works
1- Cross section through viaduct at pier centreline. 2- Cross sections through deck
•in span
and at abutment centerline: reinforcings.
CONSTRUCTION - Civil engineering Works
View of the structure. 4- General plan. 5- Form for the pour of the deck. 6- De•tail,3-showing
slack reinforcing and prestressing cables. 7- Centering to support the
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
“Roccaprebalza” viaduct
A15 Roccaprebalza road viaduct (North carriageway)
Cisa Motorway A15: upgrading of the motorway
route at the Vigne viaduct
Autocamionale della Cisa S.p.A.
Studio SINTECNA - Prof. Giuseppe Mancini CE
L.A.S. S.c.a.r.1.
Year of completion
The viaduct displays a continuous-beam scheme having a single-light caisson section of variable
depth, built cantilevered from the piers with span lengths of 120 m. Only in the end stretch
where it is expected that landslips might take place was a statically-determinate span built, its
length 45 m. On the pier shared with the continuous deck it utilized special adjustable keepers
that permit the pier to freely move without generating stress states in the decks. Design opted
for a mixed steel-concrete solution with steel webs and r.c. slabs that make possible segments
of considerable length, although only relatively light pieces need be handled (the webs), the whole
being completed with in-situ-poured top and bottom slabs. The hollow piers appear as shafts,
running from 15 m to 87 m tall, of square cross section whose dimension varies with shaft height. The minimum width lies at around 25 m from the taller pier tops, so that the piers taller
than 30 m display a trumpet profile that contributes to fining the structure, while at the same
time ensuring for the taller shafts an optimum sizing against instabilities. The section of least
dimensions displays plan dimensions of 6x6 m. For the pier and abutment foundations circular
caissons filled with concrete were used, their diameters varying from 9 m to 14 m.This solution
was dictated by the nature of the soils (fractured and scaly oxylites), by the considerable forces
unloaded by the viaduct and by the need to counter landslips in the viaduct's end area.The viaduct in question is the first structure in Italy in which the caisson deck, built of wed segments
cantileverwise from the pier, is of mixed structure, with the use of both outside prestressing and
bonded-cable prestressing.This choice enabled construction of the 9-meter-long segments owing
to the fact that the elements to be handled consisted of the steel webs only (and were thus of
relatively low weight), a building-construction crane anchored to the pier being used. Once the
webs were positioned they were connected together with a system of provisional guys and struts
that made it possible to obtain the section form desired, the distortions necessary in the stretch in tangent being imposed as well to create the screw-like behaviour of the section tied to the
changing crosswise slopes. Once the webs were blocked in definitive position the lower slab was
poured, using precast predalles as throwaway forms. In the final phase the upper slab was poured, using a metal form sliding within the caisson, precast pilaster strips being used to anchor
the upper prestressing cables. Once the tees were connected the outside continuity cables were
threaded, being deflected by means of two full diaphragms placed at the quarter-spans, and
finally they were tensioned to one-half design value. Thus the construction rises could be reduced. Only when the entire deck had been built and solidized, was the tensioning of all the outside cables completed, starting from the Parma side and ending on the La Spezia side.
CONSTRUCTION - Civil engineering Works
• 1- Cross section through deck in span (a) and at pier centerline (b).
CONSTRUCTION - Civil engineering Works
Cantilevered construction from the piers of the mixed steel-concrete struc•ture2-3-4.
deck. 5- Detail of the connector reinforcings for the precast pilaster strips.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
“Sandro Pertini” bridge upgrade
Sandro Pertini road bridge
Macerata, Italy
County of Macerata
Prof. Gian Michele Calvi CE, Matteo Moratti CE
Structural detailing
Prof. Gian Michele Calvi CE, Matteo Moratti CE
Prof. Gian Michele Calvi CE, Matteo Moratti CE
Contract management
Studio Calvi s.r.l., Pavia, Italy
General contractor
A.T.I. Rosi Giancarlo Costruzioni Srl, G.S. Costruzioni
Generali Srl, Cagnini Costruzioni Srl, Costruzioni Edili
Sirolesi Srl, Dell’Orso Perforazioni Srl
In June 2005 the masonry bridge spanning the Potenza river near the town of Macerata (92
metres long with five 18 m spans) was closed to the vehicular and pedestrian traffic because
of evidence of structural damage, apparently due to significant foundation settlement.
The importance of the bridge (average traffic 24,000 vehicle/day) required a rapid intervention,
that should have also addressed the problem of seismic safety (with an expected peak ground
acceleration of about 0.25 g for a 10% probability of exceedence in 50 years).
The architectural and historical relevance did not allow signigicant changes in the aesthetics of
the bridge.The proposed solution was to built a new concrete bridge inside the old masonry one,
only emerging with the new deck, wider than the original one to adequate the geometry to the
present prescription for vehicular and pedestrian traffic.
The final design and the work program were developed in ten days. Works were completed in
40 days.
The conceptual design was very simple:
1. two piles (D=1,2 m, length 30 m) were drilled into the three existing piers and at the abutments;
2. all material above the arch structures was removed;
3. five transverse cross beams were casted above each pile couple;
4. thirteen high damping rubber bearings were installed, 2 on each abutment and 3 on each pier;
5. a deck composed of precast extruded beams was placed on temporary supports;
6. top slab and six diaphragm beams were casted in situ;
7. the temporary supports were removed;
8. the structure was completed with a waterproof membrane, protection, finishing, asphalt, parapets, joints, deck drainage, lighting systems, and traffic signals.
CONSTRUCTION - Civil engineering Works
• 1- Longitudinal section. 2- Deck plan.
CONSTRUCTION - Civil engineering Works
1 - Drilling
2 - Water bed
3 - Original concrete foundation
4 - Siffening foundation with micropiles for masonry piers
5 - Sands, silty sands and sandy silts
6 - Gravel, gravel with sands
7 - Argillaceous silts
8 - Sands with gravel
9 - Siltstone clays with thin sandy layers
CONSTRUCTION - Civil engineering Works
Emptying of the bridge, with the heads of the piles already driven. 4- The emp•tied3-bridge.
5- Longitudinal section through the project site’s geology. 6- Longitudinal section. 7- Cross section. 8- Laying the aseismic isolations. 9- Laying the precast beams.
CONSTRUCTION - Civil engineering Works
Cable-stayed footbridge over the Frodolfo
Cable-stayed footbridge over the Frodolfo river
Bormio, (Italy)
Bormio Municipality
Structural detailing
Prof. Gian Michele Calvi CE, Dario Compagnoni CE,
Matteo Moratti CE
Prof. Gian Michele Calvi CE
Contract Management
Studio Calvi s.r.l., Pavia, Italy
General contractor
G.A.L. costruzioni, Bormio, Italy
The bridge has a single span of about 66 m passing over an important road and a river, is
curved in plan and in elevation, the deck has a variable thickness of about 300 mm and is
sustained by a single pylon, hinged at the base.The main structure has been mounted in four
days, without any temporary support, using six identical prefabricated deck sections, each one
being 12 m long, and a 35 m monolithic steel pylon to which the deck sections are anchored, by means of four cables each. The total construction cost has been approximately
500,000 €. The bridge is technologically highly innovative, light, beautifully inserted in the
environment and very cost effective. The design choices were essentially guided by the environmental constraints:
• no intermediate support was really possible, and only on the west side topography and building locations permitted a relatively easy construction of foundations; on the same side, an
underground parking under construction provided some appropriate anchoring mass;
• the beauty of the valley and the presence of an ancient stone bridge required a light structure, with minimum interference with the surroundings;
• the construction time on site needed to be reduced to a minimum, to mitigate as much as
possible traffic interruption on the main road.
It was decided to design an asymmetric steel pylon, rotated both vertically (about 7 degrees)
and horizontally, in order to optimize the force distribution.
The pylon is made of a monolithic 35 m steel pipe (812 mm external diameter) with a
second external co-axial pipe (850-1100 mm variable diameter) welded to the internal one
by six radial steel wings).
The pylon is hinged at the base and its position is essentially governed by the actual loading,
with a variable inclination. Eleven tendons (52 mm maximum diameter) restrain the pylon at
the ground.
The deck is formed by five precast high performance concrete elements supported on 10
couples of thinner cables. Each segment has the same length (12 m) and the same radius
of curvature both in plan (about 300 m) and in elevation (about 1200 m).
The in–plane curved shape of the deck is effectively reacting to horizontal loads by arching
action, whilst vertically the deck is free to rotate around a horizontal axis on the west side
and is connected to the east abutments with a double-hinged 6 m long truss that allows vertical free movements and rotations of the deck.
Pylon foundation, abutments and anchor mass for the fixed cables were constructed on site,
taking advantage of the contemporary construction of an underground parking lot.The pylon
was transported overnight in a single piece and mounted with two cranes; within the subsequent three days it was possible to position and anchor the five deck sections, prefabricated
elsewhere. During construction, a temporary connection between the deck sections was provided by steel self centring couplers, later on included in concrete injections that made the
deck fully continuous. The results of the complex nonlinear time-history simulations carried
out during the design phase were later confirmed by in-situ dynamic testing, with induced vertical displacements of ± 180 mm.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Post-tensioned concrete plinth at pylon base. 2- Temporary self centring cou•plers1- (clockwise:
plan view; section B-B; detail of the steel pin; section A-A RC transversal beam cast on site). 3- Main section of the deck (hatched zone indicates RC,
dimensions in mm). 4- Main geometry of the footbridge frontal view from South. 5Horizontal hinged bearings at the West abutment. 6- Deformed shapes under live
loads of the f. e. model (from left to right: South-East view; East view; plan view).
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Bridge over Mazzocco creek
SS16 Adriatica - Bypass of the settlements of
Montesilvano and Marina di Città Sant’Angelo – job
segment 1 – Road bridge over Mazzocco creek
Pescara, Abruzzo, Italy
Prof. M. P. Petrangeli CE, E. Cipolloni CE
Management Contractor
General contractor
A&I Della Morte (Naples) - Impresa Martella,
Pescara, Italy
Year of completion
The new stretch of the SS16 (State Road) bypass between Pescara and Montesilvano features, besides an 1800 m long tunnel, a new bridge over the Mazzocco creek, described here.
Both the contracting agency and the local authorities considered the arch bridge to be a sine
qua non for the project underway, even though the creek could have been crossed with a traditional prestressed-concrete-beam bridge, built by successive segments.
Another item that could not be ignored was that the projected works had to be compatible
with the future doubling envisaged, which would upgrade this section to the characteristics of
already-existing SS16 alternate, thus obviating future demolitions.
The construction procedures were so worked out as to not need centerings, which would have
had to stay in place for a long time since they would be in the creek bed and there was no
guarantee for the whole period of construction that the centering’s ground supports would
not settle differentially, owing to the presence on a creek bank, revealed by soil studies, of
waste material originating from the driving of an old tunnel.
Mazzocco creek was crossed on a four-span bridge a total of 140 m long. The 70 m arch
span founds on circular-section caissons 15m in diameter, of maximum depth 25 m.
Pier 1 and the abutments found instead on 1200 piles up to 30 m in length, which had to
assure, besides the required bearing capacity, a minimum strain so as not to induce drops in
the arch thrust owing to interaction with the not especially good foundation soil.
The piers, of lengths up to 14 m, are two-column frames, since deck width varies between
17.2 and 26.23 m. The decks were built with 4 precast concrete U-beams 1.6 m deep, prestressed with bonded strands, their length between 24 and 24.7 m. The first span was built
instead with 18 double T beams 1 m deep.
The 4 V beams, between piers 2 and 3, bear on the crosspiece connecting the five members
constituting the arch. Each member, its double T section varying continuously from the base
to the minimum section at the crown which is 1.20 m deep, comprises three precastings jointed in place. The central element of the three is prestressed with two post-tensioned cables
of ten 0.6” super-strands, anchored on the two ends of the precasting.
The connection of the members among themselves and to the foundation footing is effected
by prestressing bars.
The precastings were built on site to limit handling. For their launching and mounting two provisional struts 15 m tall were set up, each built with five Innocenti steel-pipe towers connected together and shearbraced, bearing on two provisional footings founded on 800 piles.
In the first phase all 5+5 low elements were launched, bearing by means of supports on the
caisson footings and on struts.
After launching by means of a 300 ton crane of the central elements, the stitches were effected and the provisional struts unloaded, this being carried out progressively by means of the
lowering of a series of large screws and sand-filled boxes placed below the provisional bearings of the precastings, the structure’s strain behaviour being constantly monitored to check
on the correspondence of the shifts with theoretical.
Thus was completed the launching of the deck beams and the subsequent pour of the slabs.
CONSTRUCTION - Civil engineering Works
• 1- Bridge plan.
CONSTRUCTION - Civil engineering Works
for the arch central precasting being struck. 3- Lower arch precasting
•being2- Forms
mounted. 4- Handling the precastings. 5- Detail of the provisional support for
the precasting on the footing. 6- Longitudinal section. 7- Central precasting launch
phase. 8- Detail of the precastings’ upper connections.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Bridge over the Sacco river
Road bridge over the Sacco river
Municipality of Sgurgola, Rome, Italy
Consorzio Pegaso S.C.AR.L.
Roma I.T.S. S.p.A.
State Railway System – High Speed lines
PROGEEST S.r.l., Prof. Arch. HC E. Siviero CE, Prof.
R. Di Marco CE
Structural engineering
Enzo Siviero, Roberto Di Marco
Enzo Siviero
Management Contractor
General contractor
Consorzio PEGASO S.C.AR.L.
The bridge over the Sacco river is one of a series of operations called for by the new highways
system setup, itself made necessary by the construction of the new high-speed railway line
between Rome and Naples.
The solution adopted is an arch bridge built using high-performance curved precast-concrete
elements. They are positioned on provisional supports and locked in place by completion
pours.To balance the thrust, at the abutments the arch ends are connected by diagonal struts
to the deck, which thus takes on the role of chain and is under tension. The moment arising
from the eccentricity of this force relative to the arch thrust is balanced by the action of active tendons placed at the deck ends, so that the forces transmitted by the structure to the soil
are practically vertical.
To eliminate continuous forms, which significantly affect the structure’s overall cost, in the construction of the arches precast curved elements were used, of high-performance concrete, placed in position by the use of provisional supports between the spans and later solidized by in
situ pours.
In order to cut construction time, a solution was adopted for the deck construction involving
precast beams completed by an in situ pour.
To balance the thrust of the arches, which could be incompatible with the soil’s mechanical
and strain characteristics, their ends at the abutments are connected by diagonal struts to
the deck. It thus takes on the statics role of a chain and is thrown into tension.
The moment given rise to through the eccentricity of this force relative to the arch thrust is
balanced by a counter couple due to the action of active tendons anchored to the deck ends
and by the corresponding reactions of the piles below the abutments, so that the forces transferred to ground are prevalently vertical.
The structure has a total length of 132 meters and comprises two arches, 56 m in span with
a 5.6 m camber, which sustain the deck connected to them at the crown and at the abutments.
The structure’s “permeability” reduces its environmental impact and interference with the flow
of water, increasing the flowrate and lowering the crosswise hydraulic thrust.
Each arch is created by the assembly of 20 x 5 precast arch segments of high-performance
concrete, having a 70 cm x 50 cm section.They are set side by side to create a structure 10
m wide, then solidized together by the pour of a concrete slab of 25 cm minimum depth.
Since the bridge centerline is skew to the river’s flowline, the precast arch segments are mounted having a mutual longitudinal slide so as to follow this geometry. In each arch’s central
zone the deck slab is directly connected to the arches by ribs of variable depth held up by
the arches. At the abutments and the pier the support consists of precast inverted-T-section
beams .
The total width of the deck is 13.10 m. It is sized for a road having two 4.75-meter lanes
and two sidewalks raised up and protected by guardrails.
The abutment and pier foundations are built on large-diameter piles (150 cm), their length
varying from 16 m at one of the abutments to 33 m at the central pier.
CONSTRUCTION - Civil engineering Works
• 1- Bridge side view.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Bridge over the Santa Caterina channel
The design solution and the arrangement in plan ensure simplicity and lightness to the whole,
in perfect harmony with the surrounding landscape and fitting with the environment. In plan,
the highway bridge has a deck in the form of a parallelogram, whose major side is 64 m long,
while the width is 12 m.
As regards durability and resistance to weathering the choice of materials assures the structure the capacity to last if properly maintained. The structure’s symbolic significance too may
be considered quite as durable. Costs are in proportion to function, to structure safety, to the
advantages that the new structure produces, to the construction phases and schedules necessary to finish it, and to the positive judgement on the view of the whole at the end.
The bridge has a basic characteristic: its centerline is skew to the river’s flowline.
This design choice gives the structure a special characteristic, creating a very interesting foreshortening in perspective and a play of light and shadow.
Its centerline forms a 36° angle with the river’s flowline, this being perceptible from some
points of view but absolutely invisible from others. The result is, at times, a forced perspective that tends to lengthen the image laterally.
In the same way, the deck and arch dimensions and the imposingness of the whole structure are legible only when compared with the human dimension, bearing witness to a careful
search for proportion. Despite the strong dimensions the eye is not especially impressed by
the beam depths, whether before or after the successive pours.
Its juxtaposition with the existing bridge is natural: after having undergone restoration the
older structure will be used as a foot bridge, while continuing to be a memory of the past of
this part of the territory.
Structurally speaking the bridge comprises the following principal reinforced-concrete elements:
- a central parallelogram-plan arch 1.00 m deep, thrown across a 39-meter chord;
- a horizontal parallelogram deck 1.00 m deep and a width perpendicular to the sides of 7.00 m;
- along the lateral edges two wings are cantilevered, each 2.50 m wide, thus bringing total
deck width to 12.00 m;
- two diagonal lateral plates, 1.00 m thick, connect the central-arch springers with the end
edges of the deck plate;
- two support walls in the riverbed, created by continuous bulkheads 1.00 m thick, having a
plan length of 12.5 m, thrust down to a depth of 22.00 m below the arch springer;
- two support walls at the banks, built from continuous bulkheads 1.00m thick, having a
length in plan of 12.50 m, thrust down to a depth of 18.00 m below the springer of the horizontal upper deck.
Road bridge over the Santa Caterina
Sant’Urbano, Padua, Italy
Municipality of Sant’Urbano, Province of Padua
PROGEEST S.r.l., Prof. HC E. Siviero arch. CE
Structural engineering
Enzo Siviero, Luigi Rebonato
Enzo Siviero
Management Contractor
Municipality of Sant’Urbano
General contractor
Impresa Locatelli, Impresa Thiene
CONSTRUCTION - Civil engineering Works
• 1- Longitudinal section: slab reinforcings. 2- Cross section through deck.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
3- Longitudinal section: foundations. 4- Plan of the highway route in which the
is fit. 5- Longitudinal section: reinforcings of the arches and of the diagonal
CONSTRUCTION - Civil engineering Works
Bridge over the Cimadolmo branch
Road bridge over the Cimadolmo branch
Cimadolmo, Treviso, Italy
Province of Treviso
PROGEEST S.r.l., Prof. Arch. HC E. Siviero CE
Structural engineering
Enzo Siviero, Luigi Rebonato, Federico Zago
Alessandro Stocco
Management Contractor
Province of Treviso – Lucio Bottan arch.
General contractor
F.lli PACCAGNAN S.p.A., Ponzano Veneto (Treviso)
Year of completion
As part of the general project for upgrading the roadway system to which S.P. 92 belongs, the
Treviso Provincial Administration indicated its intention to widen the structures so as to enable them to take the carriageway common to type V CNR roads (extraurban C2 according to
the new operating standards for highway construction, Ministerial Decree of 11 May 2001).
This involved upgrading the carriageway to a total width of 9.50 m and inserting service
sidewalks on both sides. During this phase the construction of just the deck holding the vehicle way is envisaged, while the sidewalks will have to await new financing.
The considerable use of precasting technology for the slab enabled shortening the construction-yard phase, greatly reducing the inconveniences caused to traffic. The system designed
envisaged the construction of six typological slabs combined in four types of segments to define the bridge plan for a total length of 420 m, broken down into 20 m spans.
The deck comprises three principal full-web beams 1000 mm deep, spaced 3500 mm apart,
connected to the 26 cm deep r.c. slab by Nelson-type rungs. The slab is constituted of precast plates having a 2 m module, broken down into two parts to cover the entire carriageway
width. The plate depth coincides in the central portion with the slab’s finished depth, while in
some areas (lateral relative to the plates) and in the zone where it is supported on the steel
beams, the slab is lowered to permit an in situ pour of concrete, in order to bring about static continuity both longitudinally and crosswise, as well as with the steel beams below.The plates are so positioned as to supply the crosswise slope of the roadway plane, making the depth
of the asphalt pavement constant over the entire roadway surface, and consistent with the
behaviour of the centerline in plan. Envisaged are trestlework diaphragms (crosspieces) distributed at 4.00 m spacings and connected by diagonal shearbraces placed at the level of the
lower flat arch in such fashion as to form a structure of considerable torsional stiffness, able
to distribute the eccentric loads almost uniformly over the three principal deck beams. The
total deck width is 10.70 m, of which 9.50 m are carriageway, while 60 cm for each side of
the cross section are used to hold the guardrail and its cladding. On each pier and on the
abutments elastofip-type bearings are called for, made up of a coupling of neoprene, steel
and confined teflon, so as to create unidirectional bearings for the central beams, and multidirectional ones for the border beams, able to permit the shifts due to temperature changes
or to creep. For dynamic forces of the instantaneous type (earthquake, braking) it is provided
that the bearing for the central beam be fixed longitudinally as well, so as to distribute over
each pier the dynamic forces falling to it. Called for anyway is the insertion of a fixed bearing
into one of the two central piers. Called for on the abutments is the insertion of expansion
joints able to take the shifts due to the summation of slow and dynamic forces. It has been
ascertained, on the basis of preliminary calculations and after checking the original statics
relationships, that the hollow box structure overlying the beams’ current support plane is not
essential to the strength of the pier-pulvino complex and will thus be demolished (three sides).
Therefore, the construction can be prospected, above the pulvino itself, of a continuous steelconcrete deck outfitted with expansion joints only at the abutments.The structure will appear
slenderer than the preceding one (H=1.26 m versus H=1.58 m). User comfort will be increased and maintenance costs reduced owing to the radically reduced number of joints and bearings. Under the conditions indicated above, the carriageway can be widened to 9.50 m. In
relation to traffic conditions (current and future) the placement of lateral containment barriers is envisaged, sized for an H2 impact-severity index. On the outside of the curbing the
placement of a metal cladding 2.80 m high is called for, its function to clad the deck as well
as to provide partial protection of the roadway surface.
CONSTRUCTION - Civil engineering Works
1- Axonometric view of the plate constituting the floor slab. 2- Axonometric ex•ploded
view of the floor slab. 3- The paired piers, placed with a 20 m interaxial distance. 4-5 Laying the precast slabs.
1. precast plate
2. completion pour
3. stiffening socle
4. binder
5. finish layer
CONSTRUCTION - Civil engineering Works
“Isola della Scala” bridge
Road bridge
Isola della Scala, Verona, Italy
ANAS S.p.A. (Italian Road Administration)
Original design
Redraft design and structural engineering
Prof. Enzo Siviero CE, Prof. Bruno Briseghella CE,
Prof. Tobia Zordan CE
The project presented concerns a flyover which construction was completed in 2007, located
in Verona - Isola della Scala, Italy. The total length of the structure is approximately of 400m
with 13 spans.
According to the designers’ knowledge, this is at the moment the longest IAB ever built. The
construction of the bridge, initiated in 2001 as a simply supported flyover, was interrupted
after 2 years because of economical problems. At the time of interruption, all pre-stressed concrete girders had been nevertheless purchased.
At the beginning of 2006, works restarted with a new proposal, aiming to improve the quality of the structure and change the static scheme from simply supported to integral abutment without changing the built parts, namely, the rigid abutments and the piers, in the purpose of not to increase the cost of the final structure.
During refurbishment, in order to achieve an IAB eliminating all bearings and expansion joints,
continuity was attained at the pier caps with the casting of concrete diaphragms between the
beams of adjacent spans, in order to achieve negative moment resistance. Hogging moment
resistance was also determined with a similar technique at the abutments for the end bays.
Connection between adjacent beams was carried out casting the concrete of the diaphragms
also inside the V-shaped girders for a length of 2 meters.
The bridge was opened to traffic in 2007; no mentionable damages have been noticed until
now, except for some cracks in the approach slabs.
Management Contractor
General contractor
Nuova Bitumi srl, Trento
Year of completion
CONSTRUCTION - Civil engineering Works
1- Typical cross section. 2- Longitudinal section at the pier. 3-4 Detail of the deck
5- Typical crosspiece during construction. 6- The bridge in an advanced
phase of construction. 7- Viaduct scheme.
CONSTRUCTION - Civil engineering Works
The “Strada dei Parchi”
The first most urgent operations to make safe the
structures involved in the earthquake of April 6th
Motorway A24 – Various viaducts between 99+000
km and 116+500 km
Motorway A25 - Popoli viaduct
Strada dei Parchi S.p.A.
Firm Responsible for the proceeding
Strada dei Parchi S.p.A. – Marco Carlo Rocchi CE
SPEA Ingegneria Europea – Fulvio di Taddeo CE
Supervision of construction
Strada dei Parchi S.p.A. – Luca Bartoccini CE
General contractor
Year of completion
Main features
The project calls for the provisional restoration to
function of the structures, with a view to re-opening
the following viaducts to traffic:
A24 - Fornaca viaduct,
A24 - Genzano viaduct,
A24 - Raio viaduct,
A24 - Aterno viaduct,
A24 - S.S. 17 viaduct,
A24 - Fosso Vetoio viaduct
A24 - Pettino viaduct
A24 - S. Sisto viaduct
A24 - S.Giacomo viaduct
A24 - viaduct on the L’Aquila Est interchange
A24 - Le Campane viaduct
A24 - Palude viaduct
A24 - Viadotto Vigne Basse
A24 - Costa del Molino viaduct
A25 - Popoli viaduct
On April 6th 2009 at 3.32 AM the L’Aquila area was struck by a strong earthquake (Richter magnitude (Ml) of 5.8, Moment magnitude (Mw) of 6.3, 8th-9th degree on the Mercalli scale).The earthquake sequence continued developing with a great many aftershocks, more than 20,000 of them
recorded as of June 10th 2009. 31 of them had a M1 lying between 3.5 and 5 and three had a
magnitude exceeding 5 (April 6th M1=5.8, April 7th M1=5.3, April 9th Ml=5.1)
The plan distribution of the aftershocks brings out very well the area concerned by the earthquake
sequence, extending for more than 30 km in the NW-SE direction, parallel to the axis of the
Apennine chain.
The earthquakes of the sequence took place for the most part in the upper crust, a depth of 1012 km. Only the event Ml=5.3 of April 7th to the SE of L’Aquila was as deep as 15 km.The data
gathered to date (seismicity, GPS, SAR, geology) agree in identifying the structure responsible for the
main shock as a fault having direct movement that extends some 15 km in the NW-SE direction
with a SW dip. Its extension on the surface is located in correspondence with the Paganica fault.
The damage in the epicenter zone was due, not only to the size of the earthquake (and therefore to its magnitude), but also to the break’s direction of propagation and to the soil geology.
In particular, the major damage is observed in the direction along which the faulting propagates
(effect of the source directivity) and is amplified in the areas where “soft” sediments (such as alluvial deposits, earth fill, etc.) lie on the surface.
In the case of the L’Aquila earthquake, the break associated with the April 6th event was propagated from below upwards (and therefore towards the city of L’Aquila) and from northwest to
southeast, towards the Aterno valley.
Motorways A24 and A25 have a total length of 281.4 km, and feature 174 viaducts of various
length and typology.Their total length sums to 58.3 km (21% of highway length).They were built
between the end of the sixties and the first half of the eighties. Added to these are 77 overcrossings so that there is a high incidence of crossing structures, owing to the territory’s orography.
The seismic event involved an extensive area crossed by motorways A24 and A25, causing much
damage to the infrastructures.
After the shock that struck at 3.32AM of April 6th 2009, Strada dei Parchi SpA activated its own
engineering structures and those of the principal companies in the sector to carry out checks on
the infrastructures managed.
The most serious damage observed was that done to the bearings of forty spans on nine viaducts.
The consequent discontinuities at the expansion joints exhibited 10-20 cm steps.
The damage undergone by the roadway infrastructures varied depending on their zone. Three
motorway stretches (two on A24 and one on A25) can in fact be identified exhibiting very obvious
• A24 –from the Tornimparte interchange to that of L’Aquila West, featuring moderate damage to
viaducts and settlements of embankments at the viaduct abutments;
• A24 –from the L’Aquila West interchange to that of Assergi, featuring both important damage to
the viaducts (particularly to the S.Sisto viaduct), and breakage of the motorway embankment and
its settlements at the abutments;
• A25 –from the exit for Pratola Peligna to the exit for Bussi, with motorway embankment settlements at the abutments of some viaducts and serious damage to the Popoli viaduct.
The activities of monitoring the structures started up by Strada dei Parchi SpA and the very first
repairs carried out right from the start of the event kept the motorway open for rescue vehicles.
On the other hand, the activities necessary to immediately restore the infrastructures enabled reopening to traffic the stretch lying between the Tornimparte exit and that of L’Aquila West at 8.00PM
of April 6th, without any limitations on speed. On this stretch Civil Defence anyway kept in force the
limitation for vehicles heavier than 7.5 tons in order to handle the rescue vehicle flows.
CONSTRUCTION - Civil engineering Works
1- Map of L’Aquila area, showing active faults. 2- Raio viaduct: yielding of the
The motorway stretch between the L’Aquila
West station and that of L’Aquila East was
instead reopened to traffic on Friday April
10th at 5.50 PM on the S.Sisto viaduct.This
involved using one two-lane carriageway with
traffic in both directions, but without weight
limits and with the sole constraint that top
speed be limited to 60 kmph. For the left
carriageway the situation was instead more
serious and more restoration time was needed.
On A25, in the stretch lying between the station of Pratola Peligna and the Bussi station,
the critical element was the Popoli viaduct,
with separate carriageways. In particular the
left carriageway (Pescara to Rome) was
made transitable on an emergency basis
only for light vehicles right from the first
minutes after the earthquake.Thanks to the
restoration operations carried out, on Thursday afternoon of April 9th all the weight constraints that conditioned transitability could
be removed. Traffic then ran in both senses
of flow on a single carriageway up through
June 5th 2009, when the detour was removed and circulation returned to normal.
In all these operations, contracted to Toto
SpA through the Extreme Urgency procedure, a daily average number of eighty persons
were working with the numerous equipment
items available (by-bridges, cranes, hoisting
equipment, trucks, cutters, finishing machines, rollers, etc.) as well as seventy persons
between engineers and workmen of Strada
dei Parchi SpA.
To be highlighted in particular is the uncommon spirit of self-sacrifice and emotional
involvement exhibited by the executives,
engineers and workmen involved in the job,
who worked under conditions truly at the
limit. Only thanks to this spirit of sacrifice
was it possible to assure full-capacity driving
along the motorway on April 10th.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
involved in the earthquake of April 6 2009. 4- San Sisto viaduct: breaka•ge 3-ofAreas
the roller and detachment of rack from the rack-roller supports. 5- San Sisto
viaduct: step resulting from breakage of the supports. 6- Vigne Basse viaduct: yielding of the embankment on the abutment body. 7- Popoli viaduct: step on the
platform owing to the expulsion of the roller from its support. 8- Le Campane viaduct: complete expulsion of the roller from the roller-rack support. 9- Vigne Basse
viaduct: disarrangement of the rack-and-roller support. 10- Popoli viaduct: expulsion
of the roller of the rack-and-roller support at the abutment. 11- Popoli viaduct: expulsion of the rollers of the rack-and-roller supports.
CONSTRUCTION - Civil engineering Works
A24 – Completion of the motorway
A24 – Completion of the motorway
Motorway A24 - Villa Vomano-Teramo stretch
Strada dei Parchi S.p.A.
Firm Responsible for the proceeding
Strada dei Parchi S.p.A. – Marco Carlo Rocchi CE
TOTO S.p.A. – Vincenzo Consalvo CE
Supervision of construction
Strada dei Parchi S.p.A. – Ernesto Maffei CE
General contractor
Year of completion
The job involved the completion of motorway A24 in the Villa Vomano-Teramo stretch, by the reconstruction of the east carriageway for a length of 5.7 km. The principal structures included in the
project are:
• Vomano viaduct: a structure of 770 m long, on 44 m spans whose first two spans on the
Rome side run on a continuous beam having a mixed steel-concrete structure.The remaining part
runs on simply-supported prestressed-concrete decks having spans from 26.70 m to 35.40 m, and
a continuous-slab deck.The viaduct displays over its entire length a left curve in plan.The roadway
platform of the new east carriageway is a total 13.00 m wide, with two curbings of 65 cm and
115 cm, and a carriageway 11.20 m wide.
• S. Antonio viaduct: a structure having a total length of 2500 m with simply-supported beams
and 33.80 m and 35.30 m spans, with a continuous slab in 500-meter-long segments, which subdivides the viaduct into five stretches.The roadway platform is identical to that described above for
the Vomano viaduct.
• Carestia Tunnel: the structure has a total length of 824 m with a route in plan and a grade
that follows the existing adjacent west-carriageway tunnel of the same name.The soils involved in
the driving are Miocene marls, mantled by layers of detritus.
The stretch in driven tunnel is 731 m long, while at the tunnel mouths there are two stretches in
cut-and-cover tunnel 49 m long on the L’Aquila side and 33 m long on the Teramo side. The distance between the two tunnel centerlines is 40 meters.
The average section is 173 square meters with a driving width of 15.85 m and a height of 13.40
m, and a tunnel soffit profile having a radius of 6.77 m.The tunnel roof reaches peak heights of
60 meters above the soffit in the central stretch, while it stays rather moderate (10.20 m) at the
two mouths for a total stretch of 300 m. At spacings of 300 m are two emergency foot by-passes.The tunnel was outfitted with modern safety systems.
CONSTRUCTION - Civil engineering Works
A24 Towards L’Aquila
1- A24: Panoramic view of the Vomano viaduct. 2- A24: Panoramic view of the S.
viaduct. 3- A24: Mouth of the new barrel of the Carestia tunnel.
1.Vomano viaduct
2. S. Antonio viaduct
3. Carestia tunnel
4. Underpass transitable by car
5. Subway
6. Interchange underpass
7. Embankment
8. Abutment wall (BVV)
9. Beams field.
10. Begin job segment.
A24 Towards Teramo
CONSTRUCTION - Civil engineering Works
“S. Antonio” viaduct
Motorway A24 – Rome-L’Aquila Teramo –
Completion of the doubling between Villa Vomano
and Teramo – the new S. Antonio viaduct
Villa Vomano, Teramo, Abruzzo, Italy
Strada Dei Parchi S.p.a
Holder of the Motorway A24 Rome-L’Aquila Teramo
concession for ANAS
Prof. M. P. Petrangeli CE
Management Contractor
Strada Dei Parchi S.p.a.
General contractor
TOTO Costruzioni S.p.A., Chieti, Italy
Year of completion
The operation is necessary to doubling the roadway in the motorway’s last stretch of remaining single-lane-per-carriageway; it runs from Villa Vomana to Teramo. The job as designed, its
overall amount being 110 million euros, consists of building all the structures needed for doubling the stretch.
Among the structures, the S. Antonio viaduct is the most important economically, but also engineeringwise and construction-wise. It comprises 72 spans 35 m in length with decks comprising two precast beams and a continuous slab poured in place. The essential aspects in
addition to those just mentioned concern the study of this viaduct’s foundations.
In the stretch in question, the infrastructure in fact runs through many landslip-prone areas.
This fact led the designers of the existing viaduct 20 years ago to use caisson foundations for
all the piers. In our case, careful study of the site geomorphology led to a differentiation of
the foundation type. In fact, for the valley bottom areas the choice fell on more economical
foundations on piles (1500).
The Decks
The decks are built of precast U beams 2.0 m deep, prestressed with bonded cables (80T15).
The beams are connected in place by the pour of the upper concrete slab 25 cm deep.
Continuity under horizontal forces is provided by the slab, which at the beam heads, owing to
expedients aimed at reducing its depth, enables transfer of the longitudinal stresses in its own
plane and hence on the one hand the use of the technique of precasting the beams on a
supported scheme (with obvious economic and construction-time advantages) and on the
other the possibility of seismically isolating the viaducts (Precast beams: 45 MPa; In situ
pours: 30 MPa; Concrete incidence: 7.5 m3/ml; Steel: slack for r.c. fy = 430 MPa; incidence:
200 kg/m3; Bounded-cable prestressing: fptk = 1860 MPa; incidence: 14.5 kg/m2; Cross prestressing bars: fptk = 1230 MPa; incidence: 1,8 kg/m2).
Seismic isolation
It was mentioned that the site’s strong seismicity (ag=0.25) and the need to limit the forces
at the foundation led to widespread use of seismic isolation of the deck from the substructures. In fact, the site’s orography and the constraints on the design grade profile (the viaduct
lies at the mouth of an existing tunnel) demanded piers of quite variable height, though
always less than 18.01 m, and hence an irregular viaduct, low ductility and high foundation stresses. Thus, seismic isolation both crosswise and longitudinally was obligatory.
The viaduct was broken down into five sub viaducts having a varying number of spans (an
average of 15), whose constraints scheme envisages, longitudinally, connection with elastomer
isolations at the five central piers and moveable constraints on all the other piers and abutments. In the crosswise direction fixed constraints on the abutments or joint piers and elastomer isolations of varying stiffness on all other piers. By acting on the isolations’ stiffnesses
(32000kN/m ÷ 80000kN/m), the forces on the foundation could be regularized and diffused practically uniformly over all piers, whatever be their height. The result was the adoption
of joint devices and most especially of slides for the moveable bearings, of significant dimensions: (l 260 mm; t 150 mm).
CONSTRUCTION - Civil engineering Works
• 1- General plan, from pier 0 to pier 45.
CONSTRUCTION - Civil engineering Works
for the foundation caissons (Ø = 12 m; h = 20 m). 3- Construc•tion2-ofExcavation
a prestressed beam. 4- The beam precasting field was located in the West
construction yard. In view is the tensioning head for the adherent-strand prestressing cables. 5- Longitudinal profile: new East carriageway works. 6- Cross section through prestressed-concrete beam. 7- Deck construction phase, with the launch of
the beams by steel launch car bearing on the piers. 8- The new viaduct: in view is
the S. Antonio ditch, repositioned and diverted between the two viaducts. 9- Precast
beam plan.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
Viaduct for the Algeria East-West Motorway
Road viaduct
Algeria, El Affroun-Hoceinia job segment
A.N.A. National Motorway Agency
Studio SINTECNA – Prof. Giuseppe Mancini CE
Cooperativa Muratori & Cementisti CMC, Ravenna
The viaducts were built as part of the construction of a stretch of 15 kilometers of the EastWest motorway in Algeria. Both were built of prestressed prefabricated reinforced-concrete
segments having a caisson cross section.
The structure at station 50.1 kilometers comprised a continuous deck of two separate carriageways, each of five spans having end spans 45.50 m long and intermediate spans 70 or
80 m long. Both carriageways are in curve, its plan radius 500 m.
The structure at station 49.2 kilometers also comprises a continuous deck of two separate
carriageways. It has eight or nine spans, each having end spans of 35 m or 45 m and intermediate spans of 50, 60, 70 or 80 meters. Both carriageways are in curve, its plan radius
500 m.
The 15.57 m wide individual carriageway was built having a single-light caisson section, its
depth varying from 2.80 m (in midspan) to 4.00 m (at the piertop). Both viaducts were seismically isolated using elastoplastic dissipators both longitudinally, at one abutment, and crosswise (at the piers).The degree of seismicity corresponds to a peak acceleration at the ground
of 0.35 g. The foundations are on 1200 mm diameter piles. The construction system called
for precast segments using the “short line” system, the segment earlier cast being used as
form wall for the next one. Launching was effected by a launch car, the segments were mounted with provisional prestressing bars; the tee prestressing cables being subsequently tensioned.
Year of completion
CONSTRUCTION - Civil engineering Works
1-. Cross section through double-deck viaduct. 2- Structural steel for the span
CONSTRUCTION - Civil engineering Works
and installation of the deck precast segments. 5- Structural steel
•for3-4the Handling
pierhead segment. 6- Steel reinforcing of the segment. 7- Dissipator for the
structure’s seismic isolation.
CONSTRUCTION - Civil engineering Works
CONSTRUCTION - Civil engineering Works
MAXXI – Center for the contemporary
Winner A.I.C.A.P. Award 2009 for Structural Concrete Works - Category “Buildings”
MAXXI – National Museum of Arts of the XXI Century
Rome, Italy
Ministry for Cultural Assets and Activities – Department
for architecture and contemporary art
Ministry of Infrastructure and Transport – Public Works
Superintendence of Regione Lazio: Angelo Balducci R.U.P./Proceedings’ person in charge: Roberto Linetti –
Construction yard management: Roberto Tartaro Architectural aspects’ operating director: Mario Avagnina
Architectural design
Designers: Zaha Hadid – Patrik Schumacher
Project leader: Gianluca Racana (Zaha Hadid Limited)
Structural consultants
Anthony Hunt Ass., OK Design Group
In Rome’s Flaminio district, nearby the capital’s new Auditorium, has risen the MAXXI, the
national museum for the arts of the 21st (XXI) century, conceived by architect Zaha Hadid.
Its construction began in 2003, in a construction yard of experiment and innovation.The new
structure, which houses museums and cultural activities as well as workshops and exposition
spaces, brought about a great transformation of the entire block. And this even if the design
solution adopted took its point of departure from a reading of the context, configuring a fabric
and a volume in continuity with the strictly horizontal lines of the surroundings. Entrance to
the MAXXI is gained in the heart of the block.
From the building-high hall access is had to two museums – Maxxi Art and Maxxi architecture – and to the reception services, the cafeteria, the book shop and the spaces for temporary expositions. Outside, a foot route insinuated below the overhanging volumes follows the
building plan, restoring an urban connection interrupted by the earlier military installation
occupying the lot. The architectural and structural elements connoting the oeuvre basically
number two: the walls that delimit the exposition galleries and that determine the interlacing
of the volumes; and the transparent roof that naturally lights the rooms.
The concrete wall is the element organizing space, while the roof system is the highly innovatory technological and systems element. In fact, integrated into the roof are the skylightframe elements, the devices for controlling natural lighting, the artificial lighting fixtures, and
mechanisms for limiting heat from solar irradiation.The roof system comprises a dual glazing
and is protected on the outside by a metalgrille sunshade that, besides screening light, acts
as walkways for maintenance purposes. Cement concrete (self-compacting concrete, SCC) is
the MAXXI’s true protagonist. In fact of r.c. are the walls characterizing its form and structure, as too are the horizontal surfaces and the roof blades, entirely clad with fibre-reinforced
concrete (GRC). Concrete also forms a large part of the finishings, such as surfaces in view,
floorings and furnishings.
Final construction design
Structural design: Studio S.P.C. S.r.l. Giorgio Croci –
Aymen Herzalla
Geotechnical consulting: V. M. Santoro
Specialist consulting: A. Viskovic, S. Di Cintio, M. Francini
Assistants for the structural design: F. Croci, S. Di Carlo,
I. De Rossi, A. Bozzetti, C. Russo.
Steel staircase and “monocoque” floor design: Studio
E.D. In. s.r.l. - Fabio Brancaleoni, Marcello Colasanti
SCC mix design consultant: Mario Collepardi
Structural design validator: Antonio Maffey
Syndicate: MAXXI 2006
(Group Navarra)
Assignor: S.A.C. Società Appalti Costruzioni S.p.A.
(Group Cerasi)
Prime Contractor: Marco Odoardi
Construction site engineering manager: Roberto Rossi
Construction yard chief: Gianni Scenna
Assistant construction yard chief: Luigi Carducci
Engineering office: Daniele Centurioni, Silvia La
Pergola, Fabio Ceci
Design coordination, execution of systems: Claudio
Accounting Office: Roberto Cascino, Enrico Bottacchiari
Iwan Baan
• 1- Ground-floor plan. 2- Section through building.
“Olympic Palavela”
The renewal of “Palavela”
Turin, Italy
Agenzia Torino 2006
Final design
ing. Arnaldo De Bernardi, arch. Gae Aulenti,
ing. G. Siniscalco, arch. C. Roluti, arch. S. Basso,
arch. M. Filippi, ing. G.C. Gramoni, arch. F. Quadri,
ing. W. Peisino, ing. G. Forte, ing. E. Rosati
Construction design
ing. Valerio Actis Grosso (Project Manager), ing.
Giovanni Vallino Costassa (Structural engineer)
Yard management
prof. ing. Giorgio Siniscalco (SI.ME.TE Snc)
General contractor
A.T.I. (Maire Engineering SpA, Impresa Costruzioni
Rosso Geom. Francesco & Figli SpA, Keltermica
The affairs of the Palavela (“sail hall”) go back to 1958. In that year, as part of the Italia 61
exposition (organized in the Piedmontese capital to celebrate the centennial of Italy’s unity), a
competition was promoted for the design of a building to house the Fashion and Costumes
event.The winning solution was a smooth reinforced-concrete box vault, bearing on three points
and built using procedures like those used in the construction of the Centre des nouvelles
industries et technologies, built between 1956 and 1958 in the Défense quarter. The roof
has a hexagonal plan inscribed in a circle 150 m in diameter, and consists of a self-bearing
reinforced – and prestressed-concrete shell.The height at the crown of the arches is 29.00 m,
while the composite vault roofs an area of 14,625 square metres; the volume enclosed is
332,000 cubic metres.The vault structure consists of two slabs, each 60 mm deep, developing
over the entire roof and connected together by continuous longitudinal and crosswise ribs; its
1.30 m gross depth includes a transitable interspace 1.18 m high.The continuous monitorings
the structure was subjected to over the years gave results so encouraging as to include the building in the list of works potentially useable for the competition outfittings of the 2006 Winter
Olympics. In fact, its restructuring aimed at the creation of plant for artistic ice-skating and for
short-track. The new building, wrapped by its original roof, comprises two bodies set close
together, with a reticular steel roof that, although at different elevations, connects them together.
The southeast-southwest body is assigned to spectators of sectors 1 and 2 (7196 seats), and
the northeast-northwest body is assigned to the “Olympic family”, to the athletes and to the
media (1062 seats) for a total of 8258 seats. Structurally, the two building bodies are supported by parallel septums, on which the floor structures are made to bear as are the precast seating tiers, and by the perimetral walls. The bearing septums were built of reinforced concrete
with a fair-face finish.
Self Compacting Concrete (SCC)
Unical SpA - Gruppo BuzziUnicem
Doka Italia SpA
Images from the files (1958-1961) of phases in the construction of the Pa•lavela1-2-(“Sail
palace”). 3- The sail roof was given a preliminary restoration: one of no
particular significance since careful monitoring over the years had brought out that
the building’s structural integrity and its surfaces had remained very nearly perfect.
of the new building’s structure were created with fair-face con•crete,4- Thethe elements
structure having to perform an architectural and aesthetic task as well.
Thus design opted for self-compacting mixes, capable of assuring both high mechanical performance and aesthetic beauty, besides meeting the specific pour requisites tied to so particular a construction job. 5- Design envisaged for the Ice Palace the construction of a galvanized steel reticular roof, given the necessary soundabsorbent and sound-insulation panelling. 6- The sports facility’s plan: the ground
projection of the hexagonal sail roof completely encloses the Ice Palace structure.
7- Phases in the construction, by in situ pour, of the tank that holds the ice-skating
rink. 8- Section through the building. 9- Detail of the facade of the new Ice Palace
during the final finish phases: the mix design and the special care taken in carrying
out the pours made it possible to create quality concrete surfaces, which at the same time became the sign characterizing the entire oeuvre.
New Bocconi University
“Bocconi 2000” of Bocconi University (enlargement)
Milan, Italy
Università commerciale Luigi Bocconi, Milano
Client representative
Geom. Nicolò Di Blasi
Grafton Studio, Dublino - arch. Shelley McNamara,
arch. Yvonne Farrel
Design team
Gerard Carty, Philippe O'Sullivan, Emmett Scanlon
Project Architect
Simona Castelli
The new building housing Milan’s Luigi Bocconi Business University, officially opened on October
31st last, completes the historic campus – conceived by Giuseppe Pagano a little more than
seventy years ago and enlarged to a design by Giovanni Muzio and Ignazio Gardella – and
becomes the whole university complex’s new “entrance gate”. The construction stands at the
intersection of Viale Bligny and Via Roentgen on a 60 m x 150 m lot. Its functional program
envisages that on the 68,000 square metres of walking surface a quarter of the university’s
needs be housed, among which the offices, the departments, the classrooms, the exposition
zone and the great hall seating a thousand.
Characterizing the oeuvre too is the complex structural design worked out: the sophisticated
foundations, which reach down more than 14 m below site level, the traditional continuous concrete raft 2-3 metres deep, the prestressed-concrete floor structures on the basement levels,
and the enormous wall beams, with thicknesses of 400 mm, heights up to 30 m and 24 m
spacings. To build the enlargement of Milan’s Bocconi, concrete was used to pour almost the
whole of the bearing structures, which remain in view on the interiors and in some parts on
the outside too. In particular, for the wall beams, the bearing septums, the roof beams and the
great hall’s roof structure, self-compacting concrete (SCC) was used, a material of long-lasting
workability that ensures greater compactness of the pours and therefore better quality and a
more uniform fair face in terms of aspect and colour, as well as an improvement in the oeuvre’s mechanical strength and durability.
Many reasons led to this construction and technological choice: first of all, the design’s architectural and structural complexity, which meant the need for thickly-laced reinforcing units that
would not have permitted the introduction of vibration equipment, the heavy stresses in the
hardest-working areas, the high ambient temperatures and the jobsite’s location right in downtown Milan.
Lennart Breternitz, Matthew Beattie, Philip
Comerford, Miriam Dunn, Andreas Degn, Ann Henry,
David Leech, John Barry Lowe, Eavan Meagher, Orla
Murphy, Aoibheann Ni Mhearainn, Kieran O'Brien,
Sterrin O’Shea, Eoghan O’Shea, Michael Pike, Anna
Ryan, Maurizio Scalera, Ansgar Staudt, Gavin
Structural design and supervision of construction
Studio Pereira, Milano - ing. Emilio Pereira,
ing Vincenzo Collina, ing. Massimo Sandrelli, ing. Silvio
On-site building supervision
Progetto CMR - Marco Ferrario, Danila Aimone,
Maurizio Cantoni, Claudio Pin
Utility systems design
Amman Progetti
Lighting consultant
Metis - Claudio Valent, Marinella Patetta
Interior design
Avenue Architects - arch. Dante Bonuccelli
Acoustic and electrical consultant
ARP Service, Paolo Molina
Fire-fighting consultant
ing. Silvestre Mistretta
General Contractor
GDM Costruzioni S.p.a.
UNICAL, Studio Pereira, A. Faresin, Redazione iiC
• 1- Floor plan. 2- Construction of the basement-levels’ r.c. bearing septums.
First-basement-floor plan. 4-5 Box elements composing the above-ground bea•ring3-structure.
6- Cross section through building. It shows the common underground
base and the breakdown of the standing portion into distinct bodies. 7- Installing
a floor structure’s slack reinforcings and its prestressing.
Milanofiori 2000 – Corporate Center
Milanofiori 2000 – Corporate Center (Segments B
and C1)
Milan, Italy
Brioschi Sviluppo Immobiliare spa
Erick van Egeraat, Rotterdam
Structural engineer and supervision of
Redesco srl, ing. Mauro E. Giuliani, ing. Gianluca Vesa,
General contractor
Unionbau srl
Coordination of final design, general
supervision of construction
Intertecno spa, Milan
The job is the first part of the Milanofiori 2000 development project. Located nearby the MilanGenoa motorway, it is also served by the subway’s line 2.
It consists of a large plate housing two floors above ground of parking and a raised plaza at the
elevation of the subway’s loading platform, i.e. at 6 m above site level.
Three nine-storey office buildings stand up from this foot plaza; on it space is found for business
activities, dining and refreshment, and sundry services.
The built area is 70,000 square meters, plus another 19,500 for site-level parking.
The structures feature an inclined lay of the plans of the standing buildings, which must be wed with
the orthogonal grid of the underlying parking structures.They are built of solid concrete plates.
The buildings’ typical grid (they feature irregular projections on their sides and large free-plan openings too) is 9 m by 6 m, arranged parallelogram-wise with a 73° minor angle.
The parking grid is square, 8.1 m on a side.
The columns, their dimensions kept limited, are typically in reinforced concrete, with a steel-concrete composite section for the buildings’ lower floors.
Right from the start of the study of possible structural alternatives to serve the buildings’ architecture with economy and speed of execution of the design, the solid-r.c.-slabs solution, cast over
industrialized formwork, was seen to be most advantageous.
Analysis of the geotechnical data and the value of the acting loads yielded the selection of an indirect foundations on r.c. piles drilled using the continuous helix technology.
Three types of piles are distinguished:
a. 80 cm diameter piles, 21.5m long, with Pnom = 2500kN;
b. 60 cm diameter piles 21.5m long, with Pnom = 1400kN;
c. 60cm diameter piles 11.5m long, with, Pnom = 875kN.
The parking area columns are of r.c., of typical circular section 50 cm in diameter. Columns of
different dimensions and types are envisaged in the standing buildings
The fire resistance requirement is R90.
The columns in the standing buildings are typically circular in form 50 to 70 cm in diameter, with
a composite steel-concrete type, which includes a steel H-section.
The outer concrete of the columns is not only requested by fire protection, but is considered as
working with the steel column to provide the necessary bearing capacity.
The deck of the first floor above ground (second parking level) and the deck at plaza elevation
consist of solid r.c. plates, one 28 cm deep having REI90 fire characteristics and the other 400
mm deep with REI180 fire characteristics.Their construction is envisaged by in situ casting over
industrialized modular forms. The plate fields’ plan dimensions vary depending on architectural
layout. The typical plate dimensions are 8.10 m x 8.10 m.
A dual grid of two-way reinforcing bars is necessary, with densely reinforced zones on the extrados at the columns and on the soffit in span at points where the fields are most highly stressed.
Prefabricated bundles of bars were used and unrolled on site in order to achieve the correct spacings, speed up the construction time and reduce the placing price. A specific shear reinforcing
against punching was included at the columns. The offices floor deck comprises solid r.c. plates
300 mm deep with REI90 fire characteristics. Its construction is envisaged by in situ casting over
industrialized modular forms. The plan dimensions of the plate fields vary depending on architectural layout and measure 9.00 m by 8.50 m or 6.00 m by 8.50 m.
The need to create terraces for the top offices floor and to maintain the facade module constant
as per the inter storey type meant that the mechanical deck and the roof had to be suspended
at the terraces. In fact, since the façade columns had to be set back without increasing the deck
depth for the top offices floor, a part of the mechanical deck and its roof had to be hung from
• 1- Cross section. 2- Longitudinal section.
3-4- Construction of the first levels: the pour of the deck plates. 5- Use of pre•fabricated
bundles of rebars. 6- East façade. 7- North façade.
“Acqua minerale San Benedetto” plant
Building for industrial plant
Paese (Treviso), Italy
Acqua Minerale San Benedetto S.p.A., Scorzè (Venice)
Giuseppe Zago CE
Structural engineering
Studio di ingegneria RS – Stefano Secchi CE, Padua
General contractor
Setten Genesio S.p.A.
The new industrial installation called for the construction of various buildings on an area of
400,000 square meters, for a roofed area of 175,000 square meters. It stands within a gravel quarry and runs parallel to the Vicenza-Treviso railway line for a kilometer.
The most extensive building complex forms a heterogeneous body of plan 700 m x 120 m,
its height varying between 12 m for buildings A1 and A2 and 30 m for the remaining area.
Buildings A1 and A2 were built with reinforced- and prestressed-concrete precastings, while
buildings A3, A3/1 and A5 have steel bearing structures. The production buildings, A4 and
A4/1, were built of post-tensioned in situ-cast concrete.
Building A4 (production division) has a rectangular plan of 121 m x 164 m. On the railway
side and on the quarry side, adjacent to the building, stand four 26-meter-tall towers, in which
are created the stairwells. The building has two floors, each of net height 8.5 m, and a total
height of 30 meters. The structural grid is 15 m x 15 m with rectangular-section columns of
section sufficient to ensure, in at least one direction, a net distance between columns of not
less than 14 m.
To build the floor structures various structural solutions were evaluated.That best able to meet
specifications and most advantageous appeared to be the in situ-cast, with a grating-type floor
structure having post-tensioned ribs.
Construction needs, together with the need to create suitable expansion joints, identified the
“ideal” module as a 30 m x 30 m floor structure constituted of four meshes 15 m on a side
each. Thus was identified the dual typology of a “bearing” floor structure, statically indeterminable, on nine columns, and of a “borne” floor structure, on just three columns in line and
borne on saddles envisaged at the sides of the adjacent floor structure. Each module is a twoway ribbed floor structure, built of in situ-cast concrete with partial prestressing.The post-tensioning was effected with both bonded cables and unbounded cables.
The definitive solution calls for the construction of: 12 “bearing” modules of 1100 square
meters, 12 “borne” modules of 820 square meters, and 12 “hybrid” modules (bearing on one
side and borne on the other) of a thousand square meters each.The floor structure is lightened with re-useable aluminum forms so arranged as to create a two-dimensional ribbed structure.
The ribs are spaced 1.50 m apart and are 28 cm wide.The upper cap is 120 mm deep and
is surface-treated with quartz powder. A further surface treatment having a modified sodium
silicate base forms a protective barrier against aggressive agents and limits shrinkage effects.
The arrangement of the prestressing cables is such as to create “principal bands” 6 m wide,
aligned to the columns and prestressed with grouted-sheath cables of six and eight 0.6”
strands. On the remaining ribs are present two unbonded single-strand cables. Further grouted cables 6 meters in length are called for in the areas of greatest negative moment and
• 1- The industrial complex’s development in plan and in elevation.
lightener forms. 3- Pour phases for a bearing module. 4- Installa•tion2-ofAluminium
the curtain walls. 5- Bearing-module plan. 6- Ribbed floor-structure soffit. 7Support system with saddles. 8- Support system with shear-connectors. 9- Support
New Sky Italia headquarters
New Sky Italia headquarters – Top Management and
Television Production
Milan Rogoredo, Italy
Milano Santa Giulia S.p.A., with the technical
direction of Silvio Bernabè CE
Byron Harford & Associates – East Sydney
Structural engineer
In situ-built structures design and supervision of
construction: MSC Associati S.r.l. (Milan) – Danilo
Campagna CE, Andrea Sangalli CE;
Precast/prefabricated structures design: Gamma
Engineering (Lecco) – Gianluigi Fregosi CE,
Riccardo Castagna CE
Byron Harford & Associates – East Sydney
Management Contractor
Colombo Costruzioni S.p.A – Lecco
General Contractor
Colombo Costruzioni S.p.A – Lecco
Year of completion
Building 1 is tied to building 2 through a connector body, building 2 in its turn being tied to
building 3 through a suspended bridge. The multifloor bridge (30.15 m in span, 16.80 m
wide) over the city street is held up by structural-steel work, consisting of two trestlework girders placed on the facade to form a resisting structure as high as the entire (three-story) facade. Regarding the structural system’s capacity to stand up to horizontal forces, this is delegated to the r.c. structures of the first- and top-level floor structures, which in their plane act as
infinitely stiff beams (plates).The facade glazings transfer, at these levels, the wind loading to
the diagonals of the main trestlework beam, which have elliptical sections (composed of two
semi-elliptical pipes) with greater inertia crosswise to the loads. Except for the stair blocks,
the structure is built of precastings. The building floor structures are precast of adherentstrand prestressed concrete, the columns are plant precast (Rck=50 MPa) in forms prepared
especially for the SKY project, with different r.c. sections for transport, assembly and removal
from the forms in the plant.Their weight was kept down to forty tons. The column was then
cast in two pieces, solidly joined together during assembly phase. The central column, its section varying with height from the base, 0.90 m x 0.90 m, is a single piece twenty meters high.
It was jointed with an element whose section varied with height, the element being 20-25 m
high. The structures were mounted using precise sequences, which permitted construction of
the structures in shorter times than is usual for traditional r.c. structures. The construction
module is based on a typical grid of 8.40 m x 8.40 m, which is adapted to the various situations, the spans reaching 18.00 m (building 1) and 16.80 m (building 2).
Building 1 – technological: rectangular-plan, of dimensions 180.5 m x 28.2 m (36,000 m2
of floor structure). Composed of: basement floor structure, ground floor and seven floor structures above ground. Total height: 39.20 m.
Building 2 – offices: trapezoidal-plan, with dimensions 103 m x 26 m (22,000 m2 of floor
structure. Composed of: basement floor structure, ground floor and nine floor structures above
ground. Total height: 47.30 m. Building 1B: connector building between buildings 1 and 2
(4400 m2 of floor structure). Composed of: basement floor structure, ground floor and six
floor structures above ground. Total height: 34.45 m. Building 3: trapezoidal-plan, dimensions
12 m x 25 m (25,700 m2 of floor structure). Composed of: basement floor structure, ground
floor and eight floor structures above ground. Total height: 44.75 m. The structures (stairwells
and elevator shafts) acting as shearbracing for the buildings under horizontal forces were built
in situ (Rck=37 MPa) and tied in a second phase to the precast structure by various pourrestart and continuity systems. The soils demanded a raft foundation on single-fluid or twofluid jet-grouting columns, whether interpenetrating or tangent to one another, having a maximum diameter of 1.90 m for the greater vertical loadings and maximum height of 11.15 m
beneath the stair wells. Owing to the high level of the water table, the raft was waterproofed
using the so-called “white tank” waterproofing system. Building 1 required specific structural
analyses as well as special design choices since it had to possess considerable stiffness (maximum allowable antenna rotation: 0.01° or 36”) under horizontal forces (reference wind
speed: 110 km/h), in order to ensure alignment of the trasmitting signals to the satellites
through the roof antennas. The displacement and rotation fields were derived by analyzing
two different structural models. The first is of a generalized type and was used to evaluate
the field of floor structure horizontal displacements at the various floors. From this the horizontal rotations of the decks (rotations with axes normal to the building floor structures) could
be deduced. The second concerns the building’s top floor structure, where the trasmitting
apparatus is installed. From this latter model the rotations at the antenna bases (rotations
with axes in the plane of the floor structures) were derived due to the wind pressures on the
trasmitting antennas.
• 1- Section through buildings. 2- Plan of the complex.
3- Construction of the stairwell core structures. 4-5 Phases in construction of the
6- Section through building 2. 7- Foundations plan. 8- Typical-floor plan.
Light Pavilion
Exhibition pavilion
Grand Hotel, Como, Italy
Meta S.p.A. (Paolo De Santis)
Design and yard management
Attilio Terragni
The Light pavilion stands under Lake Como’s evocative light, in a ten-thousand-square-meter
park between the city’s historic downtown and Cernobbio. The project’s origin was quite simple: the program in fact needed a space as open to the outside as possible, without the natural light’s disturbing the staging of the events.
To meet this need the construction’s volume was intersected by several plane surfaces, arranged arbitrarily in space in such fashion as to create a series of fragments, suggesting a process that could continue on to infinity.
The intersections thus determined have become the linear voids through which a limited and
quite definite dialogue develops, between rule and free will, between structure and light.
The hall volume unveils a continual weave of cuts crossing it vertically and horizontally, taking
on the landscape’s blue and green tints. The idea behind these cuts came out of the suggestiveness evoked by a former Fascist-headquarters building standing not far off, where a crystal slash crosses the upper part of the atrium roof, putting the mountains behind the city in
relation with the cathedral nave.
Analogously, in the pavilion the irregular network of cuts takes as its aim to reveal the power
and the all-pervasiveness of the corrosive action of the lines that have over hundreds of millennia modelled the present-day forms of the Lombard lakes landscape.
Design Team
Chiara Assanelli, Luca Mangione, Maja Leonelli
Structural design
Amis Milano – Antonio Migliacci CE, Giovanni
Franchi CE
Mechanical and electrical equipment
Amman Progetti Milano
Paolo Molina CE
Abba Srl, Treviso (Person in charge: Luca
Safety engineering
CDR, Carlo Ruckstul CE
General contractor
Mondelli Battista. Construction yard supervision:
Aldo Mondelli CE
PERMASTELISA Construction yard supervision:
Alfredo Piccoli
• 1- Pavilion plan.
• 2- Sections through pavilion. 3-4 Phases in building the structure.
Agenzia Spaziale Italiana new headquarters
Tor Vergata, Rome, Italy
Agenzia Spaziale Italiana
Structural engineer
Prof. Camillo Nuti CE
with STIN Section chief: Danilo Pierucci CE, Rome
5+1AA Alfonso Femia, Gianluca Peluffo
with Annalaura Spalla arch.
Management Contractor
Infrastructures Ministry
Interregional Office for Public Works for Latium,
Abruzzo and Sardinia
Superintendent: Giovanni Guglielmi CE
Supervision of construction: Mario Avagnina arch.
General contractor
SAC Società Appalti Costruzioni S.p.A. – Rome
Project manager: Bruno Cavallaro CE
Year of completion
The Italian Space Agency’s new headquarters stands in the Tor Vergata university district, East
of Rome’s great ring road, on a sloping lot adjacent to the engineering faculty.
The complex comprises ten buildings: the semicircular main building (A), where most of the
offices are concentrated; the rectangular auditorium and atrium buildings (B’ and B”), which
eccentrically intersect the previous one, on the centerline of the faculty of engineering – they
represent the monumental aspect and act as connection with the outside; a series of minor
buildings, for offices or services, such as the library (C), the nursery-school and offices building
(D), the bank and fitness-center building (E), the doorkeeper’s office (F), the cafeteria-kitchen
(G), the laboratories-offices (I), and the underground parking garages (L).
Structural design had as reference OPCM 3274, and thus took account of the city of Rome’s
listing in seismic zone 3. This is one of the first applications of the new seismic design criteria for a complex of considerable size. The design was sifted by the Higher Public Works
Council, as a building having a value exceeding 25 million euros.
The main building A and buildings C-D are buildings having an r.c. core and a mixed steelconcrete structure, with hanging steel columns and floor structures made up of predalles and
a 9 cm deep slab.The other buildings are substantially wholly of r.c., in some cases with floor
structures made of prestressed elements (honeycomb floor structures or omega tiles).
Retaining walls of considerable height are present.
Vertical seismic joints separate building A into three bodies. In each body there are two r.c.
cores, while the remaining structure is mixed steel-concrete. The first two bodies have each
six floors, one or two of which are below ground depending on the lay of the land. The third
body stands five floors above ground.
Buildings B” (lobby) and B’ (auditorium) stand along a line that crosses main building A, with
which they share the joint. They are almost totally built of r.c., with walls of great extent in
plan. Both are lower than building A.
The zone of intersection, the focus of the complex, where four structurally - independent buildings converge, is as high as building A and is without floor structures from the second floor to
the roof. In correspondence with the facade of the concave side, on the higher floors, the
overhead passageways run between the two bodies composing the main building, bearing on
more-or-less radial septums. The overhead ways, the first two floor structures and the concave facade have vertical joints in the middle of the atrium.
To roof this space an innovatory solution was found. It consists of a steel grid with concrete
slab resting through seismic isolators on the two r.c. cores of the two independent bodies of
the main building which delimit the atrium.
The upper part of the convex facade hangs from the isolated roof. At mid height a horizontal joint separate the facade from its lower part which sticks out from steel structure of the
roof of building B. Vertical joints separate the facade from the cores of building A on which
the roof of the atrium stands.
The latter solution, involving a structure bearing on seismic isolators, was particularly apt. In
fact, the stresses induced by independent seismic motion of the cores are moderate, the structure is simple, the support scheme is particularly suited to the mixed steel-concrete structure typology and its construction proceeded rapidly.
The connection of the mixed-steel concrete floors to the two bracing cores in each of the
bodies which compose building A was effected with special metal inserts to which the slab’s
restart reinforcing were welded. These special pieces were connected to anchorage plates
buried in the cores by welding in place.
The vertical reinforcement of the structural walls stick up from the foundations for one floor
and a half without interruption, so as to exclude overlapping in the critical zones of the bracing walls where yielding may happen.
General plan of the complex. 2- Main building (A) and Auditorium (B’). (pho•to: 1-Giuseppe
Maritati). 3- Main building (A) West facade, concave side.
Building A – “Crescent” offices
Building B –Atrium-auditorium system
Building C – Library
Building D –Nursery, offices
Building E –Bank, infirmary, gymnasium
Building F –Doorkeeper’s office
Building G –Cafeteria, kitchen
Building H – Bar
Building I – Former workshop offices
Building L –Underground parking
plan. 5- Main building (A): restart steels. 6- Building atrium (B”). 7•Main4- Foundations
building (A): body 3, mounting the mixed structure after construction of the r.c.
core. 8- Main building (A) and Auditorium (B’). 9- Main building (A) – Concave side.
10- Cafeteria building structure (G). 11- Cafeteria building (G).
“Altra Sede” for the Regione Lombardia
New administrative center for Lombardia’s regional
Milan, Italy
Architectural design
Pei Cobb Freed & Partners Architects, New York,
Caputo Partnership & Sistema Duemila, Milan
Architectural Design Supervision
Arch. Henry N. Cobb
Structural Design
Prof. Franco Mola CE – ECSD S.r.l., Milan
Construction Supervisor
Infrastrutture Lombarde S.p.a. – General Manager:
Antonio Giulio Rognoni CE
General Contractor
Consorzio Torre (Impregilo S.p.A., Techint
Infrastrutture, cmb, Cile S.p.A., Montagna Costruzioni
S.r.l., Pessina Costruzioni S.p.A., Consorzio
Cooperative Costruzioni, Sirti S.p.A.)
Leading Contractor
Impregilo S.p.A.
General Manager
Gaetano Salonia CE
Site Technical Manager
Vinicio Scerri CE
Construction Site General Manager and Safety
Guglielmo Fariello CE
Doka italia S.p.A.
Precast vertical elements and beams
CSP Prefabbricati S.p.A.
Cobiax Technologies S.r.l.
Year of completion
The ‘Altra Sede’Tower, located in the center of the city of Milan, has recently been completed. On January
22, 2010 an official ceremony marked the end of the construction phase and unveiled the new building
to the citizens of Milan, who were also called to cast their vote and decide the new name of the complex: Palazzo Lombardia.The building is currently the tallest in Italy, and one of the strongest features of
the skyline of the city for years to come.
The Tower is the new administrative centre for Lombardia’s regional government in Milan; the administrative complex also includes five lower buildings (about 40 m high, called Cores 2, 3, 4, 5 and 6), surrounding the high-rise Tower (Core 1), which, 161.30 metres tall, set an Italian record.The building’s sinuous interweaving strands recall mountains, valleys, and rivers of the Lombardia region.Their curvilinear
forms are adaptable to changing functional requirements and are receptive to the region’s evolving organizational structure. In addition to its headquarter functions, the building accommodates public amenities accessible to all.The winning architectural project for the new regional headquarters was conceived,
according to the guidelines set forward by the Adminisration, by the architecture bureaus Pei Cobb Freed
& Partners from NewYork, together with Caputo partnership and Sistema Duemila from Italy in 2003.
Given the very strict construction times (October 2006-December 2009), wise and innovative choices
were enforced as for the design of structures, entirely consisting of reinforced concrete elements.The foundation system is a 4m thick reinforced concrete slab resting on soil whose load bearing capacity was preliminarily improved by means of the jet grouting technique.The total volume of concrete is about 8.000
m3. For the lower layer, 1m thick, including most of the steel rebar, SC 30/37 self compacting concrete
was used, whereas for the upper layer, 3m thick, a high performance C 30/37 concrete was employed.
The vertical structures of the Tower consist of one 15,5x16,3m inner core hosting stairways and lifts,
whose maximum thickness of the walls is 45cm, and 22 circular columns with diameters ranging from
120cm at the bottom to 65cm at the top floors.The columns are located along two curved lines and
define a structural grid of 8,60mx6,50m. For the cast in situ slabs, 35cm thick, C 40/50 concrete was
used. Regarding the construction techniques, both for the vertical load bearing structure (cores and
columns) and for the horizontal ones (slabs), it was decided to make extensive use of industrialized systems, together with high profile state-of-the-art construction technologies, such as the self-climbing formwork employed to build the core of theTower, for which high strength concrete (class C 45/55) was used.
An hybrid steel encasing/reinforced concrete system was employed to build the columns, allowing the
columns to be cast at the same time for an height of up to three floors. Slabs consisted in pre-assembled panels including the main reinforcement and polyethylene spheres with a diameter of 27cm. In this
way, the total weight of the slab was reduced by about 25% with respect to a solid section and only additional reinforcement had to be placed, thus considerably reducing the construction time.These techniques,
coupled to the self-climbing formwork, allowed the construction of each storey of the tower in an average of 5.5 days, as opposed to the 8 days needed with the traditional casting system.The synergic use
of these technologies allowed the 38 floors of the Tower, of which those from 1 to 11 have a surface of
about 2000 m2 and those from 12 to 38 have a surface of 900m2, to be built in about eleven months.
This very remarkable reduction in construction times of the structural system of the Tower was reflected
in the global construction ending 60 days ahead of schedule.The roofing of the inner plaza, with a surface of about 4200 m2, consists of a steel truss structure made of welded and bolted S355 J2 steel tubes
on top of which a double-layer Texlon ETFE (ethylene-co-tetrafluoroethylene) film is installed.The ETFE
cushions guarantee a filtering of sunlight of about 50%. EFTE was chosen since it is a transparent plastic film, lighter and more resistant than glass, with superior insulating power and easier and less expensive to install.The ETFE film is also stable to UV rays and is not altered by environmental pollution and
weather conditions. Sustainability, environmental impact, durability and reduced management and maintenance costs were also consistently pursued, by recurring to the most effective solutions for plants, photovoltaic energy production, air conditioning, façades and ventilation, such as the double-layer façade making up a ‘thermal wall’ for the internal rooms, with an automated system of sun shades conceived to minimize heat dispersion and maximize the efficiency of the air conditioning system.
1- View of the Tower through the roofing. 2- Construction: facades. 3- Con•struction
of Core 1. 4- Cross section of the slab.
Ground floor plan. 6- Columns and slab systems. 7- Tower contruction. 8•Slab5-system.
9- Core 1: reinforcement details.
New “Sant’Anna” Hospital
New Sant’Anna Hospital
Como, Italy
Infrastrutture Lombarde and Azienda Ospedaliera
Architectural and plants design
Bortolazzi Consulting
Structural design
Prof. Franco Mola CE – ECSD S.r.l., Milan
On-site Supervisor
Francesco de Probizer CE
Construction Supervisor
Infrastrutture Lombarde S.p.A. - General Manager:
Antonio Giulio Rognoni CE
General Contractor
S.A.N.C.O. SCarl (Altair, GDM Costruzioni, Aster)
Year of completion
The new Sant’Anna Hospital in Como, located in the heart of the Lombardia Region, in northern Italy, is a large hospital complex for which a short time for construction, fixed in 900 days,
was allowed by the Administration.
The complex is made of eight interconnected buildings, with heights up to 22 m, including,
respectively, a number of floors ranging between 2 and 3 floors above the ground. The underground floors range between one and three, according to the height of each building above the
ground and the ground level itself varying across the plan.The whole complex is surrounded by
reinforced concrete walls with varying height.
Two main aspects governed the design choices: the earthquake hazard for the area where the
Hospital is located and the deep foundations, which caused the foundations themselves and the
lower underground floors to be entirely below the water level.
As for the seismicity, the Lombardia Region is classified as ‘Zone N.4’ in the Italian Seismic Code,
which corresponds to a PGA of about 0.05g. Since the hospital buildings are classified as ‘strategic’, an importance coefficient of 1.4 on the PGA is also compulsory, meaning that the design
PGA value became 0.07g. Due to their limited height, the buildings are all quite rigid, so that
earthquake effects are strongly prevailing on those of wind loads: earthquake is thus the main
load condition for pre-dimensioning of the structural elements for lateral resistance (global base
The chosen structural elements were reinforced concrete cores, adequately distributed in plan,
so as to reduce torque effects due to lateral loads; as for static loading, reinforced concrete
columns were dimensioned to the vertical loads in addition to the flexural effects derived from
the lateral displacements produced by the cores and also taking into account the interaction
with slab elements under vertical loading.
The foundation system includes reinforced concrete basements for the columns, supported by
concrete piles with driven steel formwork, having a maximum length of 26m.
The cores are supported by concrete plates resting on reinforced concrete diaphragms.The plan
configuration of the buildings, with a complex distribution of columns, not lined into orthogonal
patterns, gave way to varying spans for the slab structures.The lack of regularity in the distribution of vertical elements is not very strong, though, so, except for some peculiar areas, the mean
spans for the slab were 7.50mx7.50m.The ratio between the two spans calls for elements with
a two-dimensional behavior.
The irregularities in the grids called for the use of construction systems that are totally or mostly independent on the morphology of the slabs themselves. The most competitive technique at
this regard is the use of in situ casting: continuous prestressed concrete cast-in-situ slabs resting
on the columns turn out to be the best choice. An adequate number of prestressed unbonded
steel strands was introduced in the slab, which allowed a reduction of the slab thickness to
26cm.The quantity of prestressing strands and additional rebar is globally reduced with respect
to ordinary rebar, so that the time to arrange the steel before casting is significantly shortened.
Even if the arrangement of prestressing steel must be accurately carried out and supervised,
and special construction methods to guarantee the effectiveness of prestressing must be
enforced – in particular provisional gaps, sliding supports along the cores and punching shear
additional reinforcement – construction time can still be reduced, because the formworks can
be removed as soon as after only 36 hours from casting, when concrete, whose mix-design was
accurately studied, reaches a strength of about 30MPa, implying an elastic modulus of about
3x104 MPa. In this way, a good productivity level was guaranteed, coupled to high structural performance and strong standardization in the construction process.
Overview of the completed Hospital. 2- Plan view of the complex (architect•ural1-drawing).
• 3- Column-slab joint. 4- Floor plan. 5- Vertical section. 6- Facades assembling.
“Verdi” Theatre
New “Verdi” Theatre
Pordenone, Italy
Municipality of Pordenone
Ing. Carlo Filipuzzi, arch. Paola Moretti – Interstudio
s.r.l., Udine
Structural engineer consultant
Ing. Carlo Filipuzzi
Yard management
Arch. Ermanno Dell’Agnolo
General contractor
Mazzi Impresa generale SpA, Verona
Interstudio s.r.l.
The theatre stands on a long narrow lot bounded by the orthogonal street grid, and is in a
place of transition between the historic downtown and an urban outskirts almost a large
city’s. The position of the architectural complex has not compromised the language of continuity of the sur-rounding buildings, indeed, it has relaunched their vitality through the great
appeal it enjoys within the city.The theatre main entrance opens on two broad streets: the via
Battisti and the via Martel-li, which meet at an angle, right at the building entrance and on
the entrance to the plaza.Volume-wise, three elements identify the theatre’s principal use assignments: the halls and the foyer, the stage tower, and the dressing rooms. The project called
for three halls: a principal hall seating 998 between the orchestra and three orders of balconies; a smaller one containing 160 seats, and a multi-use hall seating one hundred at the
third level, useable both for small entertainments and for rehearsals, since its dimensions are
those of the stage. The choice of theatre-hall conformation fell on the traditional horseshoe,
to guarantee good visibility and acoustics. The public accesses it through the building-high
foyer: two grand stairways serve the three balconies and a route along the perimeter of the
foyer joins the three balconies with three rings. The dressing rooms are divided into four singles, three doubles, three quadruples and six group dressing rooms. In the basement there is
only the wardrobe. The first floor of the backstage zone is taken up by some of the dressing
rooms and by the dressmaker’s. The second floor holds more dressing rooms, while the third
floor, besides other dressing rooms, also holds offices, including the press office, and in the central part the choir rehearsals hall, useable too as a conference room. The fourth floor holds
the final dressing rooms, the rest of the offices, a workshop, the carpenter’s shop (directly connected to the stage tower), and the rehearsal room.
of the project, right downtown. 2- The reinforced-concrete sculpture pla•ced1-inPlan
the foyer. 3- The foyer, characterized by the sculpture and the fair-face bearing structures of the roof and of the galleries accessing the three higher balconies.
plan at the level of the main hall orchestra seating. 5- The building’s
(level of the boxes, 12.90 m). 6- Longitudinal section through building. 7Facade, on the via Martelli.
Banca Lombarda Center
Banca Lombarda Center
Brescia, Italy
SBIM – Società Immobiliare Mobiliare S.p.A. , Brescia
Gregotti Associati International s.r.l., Milano
Structural engineer
Sajni e Zambetti s.r.l., Milano
System design
Amman Progetti S.p.A., Milano
General contractor
Colombo Costruzioni S.p.A., Lecco
Donato Di Bello – Gregotti Associati International
The construction of the Lombarda bank’s new headquarters in Brescia, in the area of urban
expansion, is the result of a design search for an alternative to the tower typology, often used
for the design of buildings having a high services concentration. The complex is conceived as
a sort of virtual cube, 50 metres on a side, consisting of two lateral wings connected on the
north by a suspended volume and on the south by a low body having truncated-pyramidal
roof and by a glazed walkway at the tenth floor. The two lateral wings are assigned to operations offices; the block to the north is reserved to halls and offices for meeting the public,
and the volume that concludes in a truncated pyramid houses the two great congresses halls,
seating 500 and 150 persons. They may be accessed independently for any public use. The
complex features a sharp contrast between the white of the marble cladding of the stair
bodies’ four corner volumes and the transparent surfaces of the continuous glazed facades.
In the east and west wings completely transparent bands of glazing alternate with glass
blocks, while for the suspended building body a “dual skin” solution was adopted with a brisesoleil towards the south, and a wholly glazed façade to the north.The corporate offices building runs up twelve floors above ground, reaching a height of 54 metres, and has two basement floors, one outfitted for services and the other for parking, the total net area being
27,000 square metres. To this is added the 25,000 square metres of underground parking
adjacent. The office building structure features two trios of steel latticework beams placed at
elevations 23.35 and 44.15 metres. Their span is 36 m and their function is to sustain the
nine-floor suspended body.The beams bear on reinforced-concrete columns stabilized by two
of the four stairwells. For the foundations, considering the soil’s low compressibility and its bearing capacity, separate direct footings were chosen.
1- Plan. 2-3 Phases in the construction of the structures: all reinforced-concrete
were in situ poured.
• 4- Ground-floor plan. 5- Second-level plan.
Boglietti Palace
Boglietti Palace – Cultural Center
Biella, Italy
Giovanni Boglietti – “Obiettivo Domani” Cultural
Arch. Alberto Rizzi, Biella
Luca Gibello, Filippo Chiocchetti, Paolo Strobino,
Francesca Frigato
Structural engineer
Orio Delpiano, A.I.R.E., Biella
General contractor
Lasimon S.a.s., Biella
Archivio Studio Rizzi, Costantino Merlini, Davide
Lovatti, Roberto Marchisotti
The building, located on the southwest outskirts of the city of Biella, is essentially characterized
by the juxtaposition of two truncated pyramids: one, upright, takes in and configures the basic
structural system, i.e. the great reinforced-concrete spherical caps representative of the entire
spatial order. The other, inverted and differently oriented relative to the axes of the system of
caps themselves, houses the main stairwell, making possible the natural statics equilibrium of
the whole. The construction develops on the whole on four levels above ground, besides the
basement floor and the roof terrace, for a total net area of 1500 square metres. Between the
ground floor, housing the accesses and reception desk, and the first floor is a mezzanine floor
having as its basic function to act as an intermediate between the spaces, all wholly devoted
to housing, small and large events, such as showings, reviews and exhibitions and the video
projections they involve. The first floor develops within the great spherical caps.
On the second and top floors, directly connected with the overlying terrace, space is found for
a cafeteria and a small multimedia library, while on the basement floor is the great exhibition
area, naturally lighted by special large skylights “excavated” in the structure of the hanging garden.This permits its conversion, as needed, into a sizeable conference hall.The building is strongly characterized by the massive use, not only of r.c., but also of marble and stone.The whole
main construction above ground bears on five columns, placed at the sides of an equilateral
triangle, that is in respect of the primary design of a truncated pyramid, which is what, substantially, the building configures. The two columns near the top of this triangle, towards the
east, are 0.60 m in diameter, while the remaining three, placed at the corners and in midspan
of the west side, are 0.70 m in diameter. Starting from the upper level (ground floor) the
columns change typology, being converted into characteristic oblique pilasters, two of which –
along the west axis – display a variable rhomboidal section owing to their tapering towards the
1- The building in plan and volume. 2- Structural system of the large spherical
3- Basement-floor plan. 4- Ground-floor plan.
5- First-floor plan. 6- View of building from the south during final finish phases.
•7- Section
through building.
“Cuore immacolato di Maria” parish
“Cuore immacolato di Maria” parish complex
Formia, Latina, Italy
Parroco Maccioni Don Gesuino – Arcivescovo di
Gaeta, Mons. Vincenzo Farano
Arch. Bernardo Re
Geothecnical advisory
Ing. Giovanni Gambacorta
Arch. Bernardo Re
The Immaculate Heart of Maria parish complex stands in an area donated by the bordering
Salesian Sacred Heart hospice. During design it seemed fitting to exploit the lot’s orographic
conformation in relation to the main street, as did the need to lower the foundation plane to
reach a soil of consistency adequate to permit direct foundations.Thus was permitted: the shifting of the catechism classrooms and the services below the sacristy into a band as long as
the lot, lighted by a light shaft with plantings; the location below the church of the multi-use
community salon seating 450; and the construction of the weekday chapel for seventy faithful. On the upper floor were located the assembly hall, with access from the sacristy, the sacristy itself with parish office, and hygienic facilities for the handicapped.
Over the sacristy, suspended from the roof, are the priest’s quarters. The bell tower, placed
pseudobaricentrically, completes the whole; it is accessed from the sacristy. The entire oeuvre
was built of fair-face white lightweight structural cement concrete, having as well an excellent
heat-transfer coefficient, which meant the adoption of walls only 250 mm thick that met
energy-consumption standards. The lightness of the statics scheme was made possible by the
special technologies applied for construction.
One such was that used for the ribbed triangular lacunar plates with fine triangular mesh that
permitted very low structural weights and were aesthetically pleasing as well as acoustically
suitable. Stimulating too was the design and construction of the five bombé “sail” walls of trapezoidal form, all different the one from the other. Extremely complex was the construction of
the belltower, especially at the points where it was grafted onto the arches and at the floor
structures where the in-depth beams arrive with strong traction forces.The sacristy floor structure is a dual slab lightened with polystyrol pads, poured in situ without break. The three St.
Andrew’s crosses, placed on each end of the frames connecting the walls to take up horizontal forces, were built on site, positioned and solidly joined in the pour of the ends of the rectangular section walls.
• 1- Church plan.
front facade. 3- The soffit of the special roof created with ribbed trian•gular2- Church
plates. 4- The wooden intermediate floor suspended from the concrete frame.
“San Giovanni Battista” parish complex
New parish complex of San Giovanni Battista
Lecce, Italy
Archdiocese of Lecce, Italy
Arch. Franco Purini, prof. Laura Thermes – Rome
Luigi Paglialunga, Massimiliano De Meo
Structural engineeer
Ing. Enzo Pierri (church), ing. Andrea Cinuzi (belltower)
Supervision of construction
Raffaele Parlangèli
General Contractor
Fratelli Marullo, Calimera (Lecce)
Studio Purini Thermes
The St. John the Baptist parish complex was built in Lecce, in the Stadio district, a zone on
the outskirts featuring numerous residential buildings and economic and subsidized public
building construction.The design of the centre was thus made up with the aim of creating an
urban pole of attraction, a true “community house”, able finally to attribute recognizability to
the entire area.
The complex comprises a series of volumes that de-limit and define a foot plaza, an internal
court and a closed space, the Walled Garden, conceived for meditation. The church’s assembly hall is square-plan, 24 m on a side. Next to it is a rectangular wing holding the sacristy
and the weekday chapel.
The assembly hall’s bearing structure is sustained by just four columns, which identify a fullheight central basin, trapezoidal in plan, connoted by a strong feeling of lightness. On the hall
perimeter are the service spaces, lower in height. The four columns are con-nected by isolated beams, at a height of eight metres. One of the beams of this internal frame extends
towards the entrance wall, piercing it and projecting beyond cantilever-wise. From it hangs the
large cross that connotes the building façade. The hall roof bearing structure, which reaches
a peak elevation of 15.65 m, is seen as a lacunar complex formed of square-section beams
1.60 m on a side. In one section, the roof cantilevers out by 7 m. In the lateral areas the
beams and the columns create a square-mesh modular grid 6 m on a side.
The frame structure arranged on the perimeter is clad with plastered tuff block walls. The
parish works building has an r.c. frame bearing structure. Of special interest is the stairway,
created by a cantilevered slab.The campanile is shaped like a portal, expanded in height.The
two vertical outside septums are 0.30 m thick, as is the septum supporting the stairway, and
have a total height of 28 m. The foundations are a one metre deep footing. The cladding for
the Parish hall and for the entrance security lock, both inside and outside, is of Lecce stone.
plan and facade (line of section 4). 2- First-floor plan and in•terior1- Ground-floor
facades (lines of sections 2 and 3).
• 3- Sezione trasversale.
“Somada” Business Center
Business Center
Residential and shopping centre “Somada”, Italy
Arch. Enzo Zacchiroli, Bologna
Precast structure design
Ing. Mauro Ferrari, Reggio Emilia
General contractor for structure
Edilquattro spa, Montecavolo di Quattro Castella (RE)
General contractor for precast structures
APE spa, Montecchio Emilia (RE)
PatriziaVirginia Belli, Bologna
The project concerns a corporate complex including offices and shops set down in and completing the residential and business center of a subdivision called Somada. The triangular-plan
building features a large internal patio-garden. Lining the patio’s three sides are the shop spaces,
on the raised ground floor and the first floor, while the spaces assigned to offices are on the 2nd
and 3th floors. At the building’s heart, besides its planted area, is the cylindrical core. Within it,
starting from the raised ground floor, the helicoidal stairs and a glazed-wall elevator develop. It
culminates in a conical-roof skylight.The shops, all enjoying extensive glazing, and the offices are
served, and accessed, on all levels by broad porticoed foot ways. On the raised ground floor, the
area fronting on the court is furnished around the whole perimeter with chairs in which to sit
and enjoy the outdoor area. Formally, the complex is characterized by a number of elements,
which spring from the search for an appropriate architectural system, filled out with the necessary functional elements.The building comprises a bearing frame structure of semi-precast typeK reinforced concrete. Being conceptually connected with the traditional reinforced-concrete
frame, it is composed of reinforced-concrete precastings (beams, stair flights) light-weight
predalles-type floor structures, prestressed-concrete honeycomb slabs and in situ-poured members (foundations, columns, perimetral and dividing walls, and the stairwells). Characteristic of
the system is the wet-type node, where beam-column structure continuity is created by the simple overlapping of reinforcings and a successive solidizing pour, made at the same time as the
floor structure is poured.The structures subsystem consists of v.r.c. semi-precastings, whose joins,
effected by reinforcings and integrating pours, create a statically indeterminate structural complex. Its performance characteristics can be traced back to an equivalent in situ-poured structure, as regards the absorption of both vertical and horizontal forces.
• 1-2 Sections through the building. 3- Second-level plan: structural details.
Established in 1959,
is the national association of the Italian cement producers
AITEC belongs to Confindustria and to CEMBUREAU, the European cement producer’s
association; its missions is to promote and to communicate the product’s technical
and economic potentials, as well as a correct image of the whole cement sector.
To this purpose AITEC shares its knowledge and information with all those who, by
profession, business or studies are interested in knowing the world of cement.
w w w . a i t e c w e b . c o m

Italian National Report